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Suggested Citation:"Chapter Two - Background." National Academies of Sciences, Engineering, and Medicine. 2005. Inspection and Management of Bridges with Fracture-Critical Details. Washington, DC: The National Academies Press. doi: 10.17226/13887.
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Suggested Citation:"Chapter Two - Background." National Academies of Sciences, Engineering, and Medicine. 2005. Inspection and Management of Bridges with Fracture-Critical Details. Washington, DC: The National Academies Press. doi: 10.17226/13887.
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Suggested Citation:"Chapter Two - Background." National Academies of Sciences, Engineering, and Medicine. 2005. Inspection and Management of Bridges with Fracture-Critical Details. Washington, DC: The National Academies Press. doi: 10.17226/13887.
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Suggested Citation:"Chapter Two - Background." National Academies of Sciences, Engineering, and Medicine. 2005. Inspection and Management of Bridges with Fracture-Critical Details. Washington, DC: The National Academies Press. doi: 10.17226/13887.
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Suggested Citation:"Chapter Two - Background." National Academies of Sciences, Engineering, and Medicine. 2005. Inspection and Management of Bridges with Fracture-Critical Details. Washington, DC: The National Academies Press. doi: 10.17226/13887.
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Suggested Citation:"Chapter Two - Background." National Academies of Sciences, Engineering, and Medicine. 2005. Inspection and Management of Bridges with Fracture-Critical Details. Washington, DC: The National Academies Press. doi: 10.17226/13887.
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Suggested Citation:"Chapter Two - Background." National Academies of Sciences, Engineering, and Medicine. 2005. Inspection and Management of Bridges with Fracture-Critical Details. Washington, DC: The National Academies Press. doi: 10.17226/13887.
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Suggested Citation:"Chapter Two - Background." National Academies of Sciences, Engineering, and Medicine. 2005. Inspection and Management of Bridges with Fracture-Critical Details. Washington, DC: The National Academies Press. doi: 10.17226/13887.
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Suggested Citation:"Chapter Two - Background." National Academies of Sciences, Engineering, and Medicine. 2005. Inspection and Management of Bridges with Fracture-Critical Details. Washington, DC: The National Academies Press. doi: 10.17226/13887.
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Suggested Citation:"Chapter Two - Background." National Academies of Sciences, Engineering, and Medicine. 2005. Inspection and Management of Bridges with Fracture-Critical Details. Washington, DC: The National Academies Press. doi: 10.17226/13887.
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Suggested Citation:"Chapter Two - Background." National Academies of Sciences, Engineering, and Medicine. 2005. Inspection and Management of Bridges with Fracture-Critical Details. Washington, DC: The National Academies Press. doi: 10.17226/13887.
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Suggested Citation:"Chapter Two - Background." National Academies of Sciences, Engineering, and Medicine. 2005. Inspection and Management of Bridges with Fracture-Critical Details. Washington, DC: The National Academies Press. doi: 10.17226/13887.
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Suggested Citation:"Chapter Two - Background." National Academies of Sciences, Engineering, and Medicine. 2005. Inspection and Management of Bridges with Fracture-Critical Details. Washington, DC: The National Academies Press. doi: 10.17226/13887.
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Suggested Citation:"Chapter Two - Background." National Academies of Sciences, Engineering, and Medicine. 2005. Inspection and Management of Bridges with Fracture-Critical Details. Washington, DC: The National Academies Press. doi: 10.17226/13887.
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Suggested Citation:"Chapter Two - Background." National Academies of Sciences, Engineering, and Medicine. 2005. Inspection and Management of Bridges with Fracture-Critical Details. Washington, DC: The National Academies Press. doi: 10.17226/13887.
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Suggested Citation:"Chapter Two - Background." National Academies of Sciences, Engineering, and Medicine. 2005. Inspection and Management of Bridges with Fracture-Critical Details. Washington, DC: The National Academies Press. doi: 10.17226/13887.
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Suggested Citation:"Chapter Two - Background." National Academies of Sciences, Engineering, and Medicine. 2005. Inspection and Management of Bridges with Fracture-Critical Details. Washington, DC: The National Academies Press. doi: 10.17226/13887.
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7DEVELOPMENT OF SPECIFICATIONS FOR STEEL BRIDGES In 1967, the Point Pleasant Bridge over the Ohio River (con- structed in 1928 and also known as the Silver Bridge) col- lapsed, resulting in 46 deaths (Figure 1). The collapse was the result of a brittle fracture of one of the nonredundant eyebars supporting the main span (5–7). As discussed later in this chapter, although there is disagreement about what should be classified as an FCM, there is no doubt that this eyebar was an FCM. There are several reasons why this catastrophe was extraordinary and is not likely to be repeated. The small flaw in the eyebar may have been caused by stress-corrosion crack- ing (SCC) (5,8), which is discussed further in Appendix A. Stress-corrosion cracking should not occur in modern bridge steel; however, in this instance, the eyebar steel was 1928 vintage, heat-treated AISI 1060 steel, which is substantially different than today’s steel. The fracture toughness of the eyebar was also marginal and a relatively small crack led to the brittle fracture of the eyebar, which in turn led to the collapse of the bridge. This collapse was the catalyst for many changes in material specifications, design, fabrication, shop inspection, in-service inspection, and maintenance of steel bridges. Material, Design, and Fabrication Specifications, and Effect of Bridge Design Date In 1974, in part as a result of the Point Pleasant Bridge col- lapse, mandatory Charpy V-notch (CVN) toughness require- ments were initiated for welds and base metal to ensure ade- quate resistance to fracture; that is, fracture toughness (9,10). The greater the CVN at a particular temperature, or the lower the temperature at which the CVN is required, the larger the critical crack size that can be tolerated at lowest anticipated service temperature without fracture. The present CVN requirements (for non-FCM) are essentially the same as the CVN requirements implemented in 1974. The CVN require- ments were the result of significant debate and some com- promise during their development, which is discussed fur- ther in Appendix A. Presently, material selection, design, and fabrication of steel bridges are governed by • AASHTO LRFD Bridge Design Specifications (1) and • AASHTO/AWS-D1.5, Bridge Welding Code (2). In addition to CVN requirements, these provisions restrict the choice of details as well as control weld flaws and other crack-like defects. These provisions have reshaped industry practices and result in an acceptably low probability of fatigue cracking and brittle fracture in new bridges. However, many older steel bridges built before the imple- mentation of modern fatigue design provisions in the mid- 1970s possess poor fatigue details, such as cover plates that can develop fatigue cracks (Figure 2) (11), which if not repaired, can grow and lead to fracture of the member and possible collapse of part or all of the bridge. Other factors that make these older bridges susceptible to fracture include: • Marginal fracture toughness of the steel and weld metal; • Detailing, fabrication quality, and shop inspection below modern standards; • Severe corrosion problems, especially at open or failing expansion joints; • Higher traffic volumes and truck weights than the bridge was originally designed to handle. In light of these factors, periodic in-service inspection is particularly important for older bridges to provide an oppor- tunity to detect cracks and corrosion before they grow to a critical size. In 1970, partly in reaction to the collapse of the CHAPTER TWO BACKGROUND FIGURE 1 Collapse of Point Pleasant Bridge.

Point Pleasant Bridge over the Ohio River, the NBIS (4) were established. Title 23, Code of Federal Regulations, Part 650, Subpart C sets forth the NBIS for all bridges of more than 20 ft (6 m) span on all public roads. Section 650.3 specifies inspection procedures and frequencies, indicates minimum qualifications for personnel, and states reporting, inventory, load posting, and inspection recordkeeping requirements. The current NBIS mandates a 2-year inspection interval for all highway bridges carrying public roads. However, modern steel bridges are not nearly as suscep- tible to fracture as older bridges (12,13). As a result, the ways modern bridges are managed could possibly be evaluated differently than older bridges. This could be studied further, with considerable potential benefits. For example, problems with severe corrosion have been reduced. In the last 20 years, durability of weathering steel and coating systems has improved. Expansion joints have been improved if not eliminated through the use of continu- ous jointless bridges (14). In addition, there have been few if any cases where weld defects or low-toughness steel has been an issue for modern steel bridges, owing primarily to improvements in details, fabrication practices, and fracture toughness of the steel and weld metal (12,15). If spontaneous fracture from weld defects is ruled out, then fracture can only occur if preceded by fatigue (15). Therefore, in this case, it is essentially sufficient to control fatigue to prevent fracture (12,15). Distortion-induced fatigue cracking, discussed further in Appendix A, continued as a fatigue problem in typical plate girder bridges designed before the mid-1980s (11,12, 15,16). A common example of distortion-induced fatigue cracking is web-gap cracking, which occurs in the gap when a connection plate is not attached to a flange and is subject to out-of-plane distortion (Figure 3). This problem was cor- 8 rected in 1985 by a change in AASHTO specifications that mandated the attachment of the connection plate to both flanges. Hence, it is important to distinguish three different age ranges of steel bridges: 1. Steel bridges built before the implementation of modern fatigue design provisions in the mid-1970s. 2. Steel bridges designed after the mid-1970s, but before 1985, which have fewer fatigue problems but remain susceptible to distortion-induced fatigue. 3. Modern steel bridges designed after 1985 that should not be susceptible to fatigue at all. Fatigue is virtually unheard of in modern steel bridges as a result of improved design specifications, more fatigue- resistant details, and improvements in shop inspection (15,16). Fatigue problems that have occurred in modern bridges were typically the result of unintended behavior or a design error that is not consistent with the intent of present specifications and usually manifest in the first few years of the life of the bridge. Although the bridges that are referred to herein as “modern” are less than 20 years old, there is substantial confidence that these structures will continue to perform with few problems resulting from fatigue. As a final note, in changing from load factor design (LFD) to LRFD, two-girder bridges will be designed more con- servatively than before. According to Dr. Dennis Mertz, in calibrating the LRFD specifications, loads were increased slightly to compensate for improved and less conservative distribution factors. However, the distribution factors for two-girder bridges were always reasonably accurate, so they did not get the benefit of improved distribution factors that multigirder bridges did. This should be kept in mind when considering the reliability of these two-girder and two-line truss systems. FIGURE 2 Development of fatigue crack at cover plate ends on the multibeam Yellow Mill Pond Bridge in Connecticut in 1976. (Courtesy: John W. Fisher.) Concrete Deck Girder Flange Girder Web Crack Cro ssfram e FIGURE 3 Typical web-gap fatigue cracking.

9Additional Material, Fabrication, and In-Service Inspection Requirements for FCMs and Cost Impact FHWA led the development of a fracture control plan (FCP) to provide a higher level of safety for FCBs. In the broad sense, an FCP includes everything that affects the potential for fracture—in-service inspection and maintenance as well as design, fabrication, and shop inspection (17). The idea is that trade-offs can be made between components of the plan without compromising reliability. For example, better tough- ness could be required to compensate for relaxed in-service inspection standards, because better toughness would lead to a larger critical crack size that would be easier to see from a distance and that would take longer to develop. Efforts to make FCBs more conservative were largely the result of experiences with cracking in tied arches, mostly owing to large fabrication defects because of difficulty associated with welding A514 steel (11). The American Iron and Steel Institute initiated a research project to develop an improved FCP for fabrication of non- redundant structures. This work (9,17–21) ultimately resulted in the 1978 publication of the AASHTO Guide Specification for Fracture Critical Non-Redundant Bridge Members (22). A key element was more stringent CVN requirements for base metal and weld metal for FCMs. This Guide Specification has now been withdrawn. The CVN requirements for base metal, including the greater requirements for FCMs, are now included in the AASHTO LRFD Bridge Design Specifications (1), as well as the ASTM and AASHTO specifications for the steel (23,24). Most of the remaining material from the Guide Specification is now included in Section 12 of AASHTO/AWS-D1.5, “AASHTO/ AWS Fracture Control Plan (FCP) for Non-Redundant Members” (2). Note that in AASHTO/AWS-D1.5 the defini- tion of an FCP is narrower, including only fabrication and shop inspection—not base metal selection, in-service inspec- tion, and maintenance. Therefore, these provisions will be referred to as a fabrication FCP. (Unfortunately, yet another completely different meaning for the term “fracture control plan” has arisen and that is the plan- or elevation-view draw- ing identifying all FCMs for use in in-service inspection.) The differences between the provisions for the fabrication of FCMs in Section 12 and the provisions for non-FCMs elsewhere in AASHTO/AWS-D1.5 are primarily more strict fabrication and shop inspection requirements to control weld flaws and other crack-like defects in FCMs. For example, transverse groove welds are required to be inspected in the shop with both radiographic testing (RT) and ultrasonic test- ing (UT), whereas only RT is required for non-FCMs. The fabrication FCP and the more stringent CVN require- ments result in an even lower probability of brittle fracture in new FCMs than for typical non-FCM members. Note that this additional fracture reliability does not apply to older FCMs designed before implementation of the FCP in 1978 (22). The survey (described in chapter three) indicated that approx- imately 75% of FCBs in present inventory were designed before the FCP. However, the fabrication provisions and the CVN require- ments for the materials of the fabrication FCP increase costs. One major bridge fabricator reported that the approximate increase in initial costs for new FCMs relative to non-FCMs is on the order of 8% of the cost of fabricated steel. In 1983, the Mianus River Bridge on I-95 in Connecticut (built in 1957) collapsed, killing three persons (Figure 4). Packout corrosion in a nonredundant pin and hanger assem- bly pushed one of the plates partly off the pin, eventually leading to a fatigue crack and collapse of the suspended span (between two cantilevers) (16,25,26). This event can be fur- ther attributed to poor maintenance, because a clogged drain was partly responsible for the packout corrosion. Through- out the country there are numerous other bridges with similar pin and hanger details; however, this type of suspended span is rarely if ever used in new designs. As with the eyebar of the Point Pleasant Bridge, this bridge collapse demonstrated that these pin and hanger assemblies are also clearly FCMs. (Also similar to the Point Pleasant Bridge collapse, the Mianus River Bridge collapse was the result of extraordinary circum- stances, in this case corrosion, and not just fatigue or fracture.) In part because of this failure, the NBIS were revised in 1988, requiring among other things a hands-on inspection of FCMs. This requirement significantly increases life-cycle costs relative to non-FCMs, which may be inspected from the ground, through, in most cases, binoculars (27–29). This requirement is particularly onerous for box girders, because it requires the inspectors to enter the boxes, which signifi- cantly increases costs. The frequency and extent of inspec- tion are not clear in the current NBIS. Consequently, there is disagreement on what constitutes a fracture-critical inspec- tion and how often it is done. (When the word inspection is FIGURE 4 Collapse of Mianus River Bridge. (Courtesy: John W. Fisher.)

used throughout the rest of this report, it is intended to mean fracture-critical, hands-on field inspection, unless otherwise noted.) These increased life-cycle costs for FCMs are signifi- cantly greater than the approximately 8% increase in initial materials and fabrication costs for FCMs discussed earlier. According to the survey (and as described later in chapter three), many owners believe that inspection costs associated with FCBs consume a large portion of their entire inspection budget. Owners were asked to estimate the relative increase in costs when inspecting an FCB relative to inspection and non-FCBs. There was substantial variation in the response; however, most agencies indicated increases of between 200% and 500%. The most common reasons indicated for these increases were: • Specialized access equipment such as a snooper (Fig- ure 5), manlift (Figure 6), or rigging required for hands- on inspection (30). • Traffic control to close lanes to permit the access equip- ment to be placed on or below the bridge (see Figure 5). • Additional employee-hours required to conduct a detailed hands-on inspection. • More frequent use of nondestructive testing (NDT) (described in Appendix A). • Greater frequency of inspection for FCBs. These hands-on inspections have revealed numerous fatigue and corrosion problems that otherwise might have escaped notice. Many of these problem details are discussed in Appen- dix A. Twenty-three percent of respondents to the survey indi- cated that they found significant cracks that could have become much worse, possibly averting collapses (see chapter three). Similar examples may be found in trade magazines [e.g., see Zettler (31)]. 10 Primarily because of these increased life-cycle costs, there is a general reluctance to design new FCBs. Fewer FCBs have been proposed since the fabrication FCP went into effect in 1978 (22). FCMs, such as steel pier caps and cross girders, are still frequently designed, although usually only if they cannot be avoided. In some circumstances, bridge designs with FCMs, such as tied arches, two-girder bridges, and trusses, may be the most efficient and cost-effective structural system. Although the more stringent CVN requirements, the fabrication FCP, and the additional inspection requirements for FCMs are beneficial, if they are overly conservative for modern bridges they can become an obstacle to the savings gained in using more cost-effective designs. International scanning tours for bridge management (32) and fabrication (33) have noted that Europe does not have special policies for FCMs. A risk-based approach, coupled with more rigorous three-dimensional analysis techniques, is used to ensure that a sufficient level of structural reliability is provided. Consequently, steel bridge designs that would be considered fracture-critical in the United States are still com- monly built without prejudice. However, they have also had failures of what we would consider FCBs. The following is from the fabrication scanning tour report (33): Perhaps the most significant design-related observation of the scan team was the rest of the industrialized world’s more liberal view of the importance of redundancy. Two-girder bridges, as well as other structure types considered nonredundant and fracture critical in the United States, are not discouraged and, in fact, are used extensively as safe and cost-effective bridge designs. Kawada Industries cited redundancy studies it performed to demonstrate adequate redundancy of its two-girder systems with widely spaced, mid-depth cross beams. No special design, fabrication, or inspection requirements for such bridges were apparent. The U.S. design philosophy for nonredundant bridges should be reconsidered, based upon these observations and improvements in steel toughness. Twin-girder railroad bridges are common in Germany. The single-cell box girder, commonly used for elevated roadways in FIGURE 5 Snooper used for hands-on inspection from bridge deck. FIGURE 6 Hands-on inspection from manlift.

11 urban areas of Japan, would be classified as fracture critical in the United States, but has provided excellent performance. Other interesting findings from the scanning tours were that the inspection frequency is risk-based in Europe and that the inspectors’ qualifications are commensurate with the complexity of the bridge. REDUNDANCY AND COLLAPSE OF STEEL BRIDGES Definition of Redundancy and Contrast to Indeterminacy and FCMs The AASHTO LRFD Bridge Design Specifications define redundancy as “the quality of a bridge that enables it to per- form its design function in the damaged state.” In NCHRP Report 406 (34), Ghosn and Moses defined superstructure redundancy as “the capability of a bridge superstructure to continue to carry loads after the damage or the failure of one of its members,” and this definition is also used in the AASHTO Manual for Condition Evaluation and Load and Resistance Factor Rating (LRFR) of Highway Bridges (3). Even though it has to do with potential performance in the event of damage, redundancy is a quality of the undamaged structure. Note that these definitions are not clear about what load type, magnitude, distribution on the bridge, dynamic ampli- fication, and load factors are supposed to be resisted by the damaged structure. Ghosn and Moses attempted to set require- ments for the residual capacity of the damaged superstruc- ture (34). The definitions of redundancy are also not clear regarding the type and extent of damage. For example, in a bolted or riveted built-up member it is likely that a fracture would be limited to only one tension element, because it cannot prop- agate directly into neighboring elements. However, a ship collision could destroy the entire member. Ultimately, it is the target level of reliability that designers and engineers who rate bridges should strive to achieve and the focus should not exclusively be on redundancy. Redun- dancy has a major impact on the risk of collapse and this impact is accounted for appropriately for all types of struc- tures in both the LRFD Specifications and the LRFR Manual, as discussed here. Using the LRFD and LRFR procedures, it is possible to achieve the target level of reliability without redundancy in a bridge that is more conservatively designed. Structural redundancy and structural indeterminacy are often confused and used interchangeably, although they are really two separate issues. Structural indeterminacy simply refers to whether or not the forces in a structure can be deter- mined with statics. A structure that is indeterminate, although possibly providing alternate load paths, would not meet the definition of redundant if it were to collapse, because the members in the alternate load path did not have sufficient capacity to carry the redistributed loads. It is also true, but less obvious, that there are determinate structures that can be shown to meet the definition of redundant by developing new alternative load paths—and even a few examples (discussed later) of determinate structures that have demonstrated that they meet the definition of redundant by surviving a signifi- cant fracture in service. As defined in the introduction, FCMs are nonredundant; however, nonredundant is a broader term because it also includes • Substructures; • Members that may be inherently not susceptible to frac- ture, such as compression members, but still could lead to collapse if damaged by overloading, earthquakes, fire, terrorism, ship or vehicle collisions; and • Members made of materials other than steel. Substructures such as piers are often nonredundant and therefore earthquakes, scour, vehicle collisions (Figure 7), and ship collisions (Figure 8) have led to most of the major collapses of both steel and concrete bridges. For example, an article published in 2002 just after the collapse of the Inter- state 40 bridge over the Arkansas River in Oklahoma, listed seven major bridge disasters in the United States up to that time (35). The two FCBs discussed previously, the Point Pleasant Bridge and the Mianus River Bridge, were 28% of the list. The remaining 72% were the result of substructure failure. • The Arkansas River Bridge, Sunshine Skyway (Florida) (1980), and the Queen Isabella Causeway (2001) bridge collapses were caused by ship collisions. • The Schoharie Creek Bridge in New York (1987) and Arroyo Pasajero Bridges in California (1995) collapses were caused by scour. FIGURE 7 Example of vehicle–bridge collision causing collapse owing to nonredundant substructure. (Courtesy: Robert Sweeney.)

In addition to prevention of collapse in the event of frac- ture, redundancy of the superstructure is important for sev- eral other reasons. The first is the need to more easily redeck the bridge. Also, as discussed earlier, events other than frac- ture can also damage and completely destroy members of the superstructure. For example, the fascia girder of the I-610 Bridge over the Houston Ship Channel was struck twice by ships, once in December 2000 and once in May 2001. The highly redundant multigirder bridge withstood each collision, although both times the bridges had to be closed for repairs. These are compelling reasons to have redundancy. (Note that periodic inspection is not really helpful in finding this type of damage from collisions or other extreme events because the damage is usually immediately obvious to the public. A deter- mination to close and repair the bridge can be made quickly 12 in these cases. This is an important point because in these cases it reflects on the maximum live load the damaged struc- ture is likely to experience in the brief period before closure. Periodic inspection may be more helpful in finding fatigue cracks and fractures because they are often not immediately obvious.) These are compelling reasons for redundancy. These reasons for redundancy (other than fracture) could be used to encourage redundancy outright instead of indi- rectly by penalizing FCMs. For example, in the LRFD Spec- ifications, redundancy is encouraged in Sections 1.3.2 and 1.3.4. Load factors are modified based on the level of redun- dancy, and it is stated that multiple-load-path and continuous structures should be used unless there are compelling reasons to do otherwise. FIGURE 8 Collapse of Queen Isabella Causeway Bridge in Texas in 2001 when pier was struck by a barge.

13 In the LRFR Manual, redundancy is reflected in system factors that reduce the capacity of each member in non- redundant systems. The system factors are calibrated so that nonredundant systems are rated more conservatively at approximately the same level of reliability associated with new bridges designed using LRFD Specifications, called the “inventory” level in former rating procedures. Redundant systems are rated at a reduced reliability level corresponding approximately to the traditional “operating” level. The sys- tem factor for the most nonredundant bridge types is 0.85. This means that a nonredundant bridge designed for 1.17 (the inverse of 0.85) times the design load has approximately equal reliability to a redundant bridge designed for 1.0 times the design load. According to Ghosn and Moses (34), a redundant super- structure has at least one alternate load path and is capable of safely supporting the specified dead loads and live loads and maintaining temporary serviceability of the deck following failure of a main load-carrying member. They recognized that redundancy is related to system behavior rather than individ- ual component behavior. The specifications generally ignore the interaction between members and structural components (i.e., system behavior) in a bridge, however. Redundancy is often discussed in terms of three types (28, 29,34): • Internal redundancy, also called member redundancy, exists when a member is comprised of multiple elements and a fracture that formed in one element cannot propa- gate directly into the adjacent elements. Examples include girders with composite deck (the deck remains intact in the event of a girder fracture as in Figure 9); riveted or bolted built-up girders, tie girders, or tension members of a truss (Figure 10); split box sections with longitudi- nal bolted splice (Figure 11); the bracing system, later- als and cross frames within a box member (Figure 12); bolted continuous plates or shapes to give a member redundancy (Figures 13–15), and single-cell concrete boxes with multiple post-tensioning strands. Note that it must be shown that the damaged member (several pos- sible cases considered separately with one element frac- tured or removed) can survive the prescribed loads to be internally redundant. • Structural redundancy is external static indeterminacy and can occur in a two or more span continuous girder or truss. Note that only part of an end span between the fracture and the pier may be supported by structural redundancy and that the part of the end span at the abut- ment could theoretically collapse. However, internal redundancy of the deck on a composite girder could be sufficient to maintain stability of the end span, espe- cially when combined with the structural redundancy, as in the Hoan Bridge shown in Figure 9. • Load-path redundancy is internal static indeterminacy arising from having three or more girders or redundant truss members. One can argue that the transverse mem- bers such as diaphragms between girders can also pro- vide load-path redundancy (see Figure 12). Note that the LRFR Manual (3) and the Bridge Inspector’s Reference Manual (29) state that “in the interest of conser- vatism” internal and structural redundancy should be neglected, meaning that load-path redundancy is the only redundancy that matters. As shown by in-service behavior of fractured FCBs discussed in the next section, neglecting all but load- path redundancy is clearly oversimplifying and possibly overconservative. Examples of Behavior of FCBs That Experienced Major Fractures Two examples of FCB collapses, the Point Pleasant Bridge (constructed in 1928, Figure 1) and the Mianus River Bridge (constructed in 1957, Figure 4), have been discussed. These are the only two examples of collapses of major steel bridges as a result of fracture in the superstructure. As explained pre- viously, there were circumstances other than just fatigue and fracture that were the root cause in both of these failures. On the other hand, there are numerous examples of bridges with members that would traditionally have been classified as FCMs that have fractured, but the bridge did not even par- tially collapse. Two-girder FCBs that have experienced either partial or full-depth fractures but did not collapse include: • The 1976 full-depth fracture of the US-52 bridge over the Mississippi River in St. Paul, Minnesota (called the Lafayette St. Bridge) (see Figure 16) (11,36). (It should be noted that during the course of this syn- thesis, it was mentioned that the bridge remained stable FIGURE 9 Example of bridge deck acting as catenary with hinge at fracture location in end span of the approach spans of the Hoan Bridge in Wisconsin—two of the three girders had full- depth fractures in December 2000.

because it leaned on an adjacent bridge). However, interviews with Donald Fleming, former Bridge Engi- neer of the Minnesota Department of Transportation (DOT) and John W. Fisher, both of whom were directly involved with the failure investigation, con- firmed that this was not the case. The bridge did sag 6.5 in. (165 mm), but was not supported by the adjacent bridge. 14 • The 1977 full-depth fracture of the I-79 bridge at Neville Island in Pittsburgh, Pennsylvania (Figure 17) (11,15). • The May 2003 fracture of the US-422 Bridge near Pottstown, Pennsylvania. The entire bottom flange and approximately 9 in. (230 mm) of the web fractured (37). It is apparent that other elements of these two-girder bridges, particularly the deck, along with the floorbeams, cross- PL 1 1/8 PL 1 3/4 PL 1 1/8 PL 1 x 7 7/8 (TYP) CL SPLICE FLANGE SPLICE CL SPLICE WEB SPLICE 3' - 0" 5'-3" 3" 6 1/ 4" 6' - 8" 6 1/ 4" 3" 2'-9" (a) (b) FIGURE 10 Examples of internally redundant members: (a) riveted built-up girder and (b) bolted built-up tie girder proposed for Blennerhassett Arch Bridge. (Courtesy: Michael Baker Jr., Inc.)

15 frames, and stringers, are sometimes able to carry the loads and prevent collapse. These alternate load paths were so robust in the I-79 and US-422 fractures that there was little or no per- ceptible deformation of the structure. For example, when a tugboat pilot discovered the I-79 fracture, the bridge was still providing a serviceable roadway. In the case of the Lafayette St. Bridge, displacements of 2.5 in. (63 mm) were noticed relative to the adjacent bridge 48 days before the fracture was discovered, growing to 6.5 in. (230 mm) as the crack length increased over that time. The less obvious nature of the damage in the case of a fracture as opposed to the other causes such as collisions discussed previously has implications for the loading to evaluate the damaged member (residual capacity). A lower level of residual capacity would be required for members damaged by these other more obvious causes because the bridge is likely to be closed within hours after the event. However, if a fracture goes unnoticed for an extended period, the probability of larger permits or illegal loads increases significantly. As explained in chapter three, some agencies even classify three-girder bridges as FCBs. The Hoan Bridge fracture, shown in Figure 9, is an example of a three-girder bridge end span (which is viewed as most critical owing to inadequate con- FIGURE 11 Splitting a box section with a bolted longitudinal splice to give it internal redundancy. (Courtesy: HNTB.) Truss action between slab and diaphragm FIGURE 12 Schematic of twin composite tub girder superstructure showing internal redundancy provided by bracing system and possible alternative load path provided by slab and diaphragm. (Courtesy: HNTB.)

16 Lower chord Redundancy plate FIGURE 13 Redundancy plate bolted to lower chord of SR-33 bridge near Easton, Pennsylvania. (Courtesy: HNTB.) FIGURE 14 Tee section bolted to continuous lower flange of box section to provide redundancy. (Courtesy: HNTB.) tinuity at the joint), with two out of three girders and the web of the third girder fractured (38). These fractures in actual bridges are the most valuable data available to judge the necessity of special provisions for FCMs. These full-scale tests are much more valuable than laboratory tests or numerical simulations, because the for- mer are subject to idealizations and assumptions. Although not sufficient to prove that two-girder bridges should not be classified as having FCMs, these incidents do show that under some circumstances they do not meet the definition of an FCM. The survey (see chapter three) revealed several other exam- ples of FCBs that had experienced major fractures but had not collapsed. These were usually noticed in an inspection, but had occurred at an earlier, unknown time. Similar accounts can be found elsewhere (31,39). Although not caused by a fracture, a train derailment on a nonredundant truss bridge, shown in Figure 18, is another example of valuable in-service behavior of a full-scale bridge. Note that several diagonals, hangers, and upper chord braces are completely severed, but the truss did not collapse even though a significant portion of the live load remained on the bridge.

17 Other interesting field tests have been performed on I-40 bridges over the Rio Grande in Albuquerque, New Mex- ico (40). The two-girder bridges, which had spans ranging from 131 to 163 ft and were classified as nonredundant fracture-critical, were built in 1963. The girders were 10 ft deep and spaced at 30 ft on center. A torch cut was used to sim- ulate a fracture of four different lengths to one of the girders in the bridge, the last of which was nearly full depth. Idriss et al. studied the redistribution of loads, the loading the bridge can withstand in the damaged condition, and the potential for col- lapse. The bridges were loaded with a truck that was 95% of New Mexico legal load and roughly equivalent to HS-18.35. Idriss et al. (40) also reported that under dead and live loads and when the truck was located above the crack, the flange only deflected 1.2 in. (28 mm). There was no sign of yield- ing and no significant change in strains experienced by the other instrumented members until the bottom flange was completely severed. This suggests that load redistribution did not occur until the bottom flange was completely severed. They also reported that most of the load was redistributed through the damaged girder and stringer deck system to the interior supports. In general, the load was redistributed from the damaged girder to the diagonal bracing, diaphragms, stringers, deck, floorbeams, and remaining girder. Reliability Studies of Redundancy A brief summary of selected research studies focusing on redundancy is presented here. More complete summaries of FIGURE 15 Retrofit redundancy plate bolted to web of existing two-girder superstructure in Poplar Street Bridge complex in East St. Louis. (Courtesy: Wiss, Janney, Elstner Associates.)

18 the individual articles and additional articles can be found in the annotated bibliography in Appendix D. A large number of studies have attempted to characterize the reliability of bridge designs with varying redundancy. In these reliability studies, the degree of redundancy of a system was examined by reviewing the difference between relia- bility indices. Ghosn and Moses (34) studied redundancy in highway bridge superstructures by examining the differ- ence between the safety indices they defined and those of bridges that have been known to perform as desired. Kritzler and Mohammadi (41) used the same approach in which they compared the safety reliability index of a redundant struc- ture considering all failure paths and the safety index of the exact same structure with no alternative load path. A reliabil- ity approach was also used by Moses (42) for the evaluation of bridge safety and remaining life. Frangopol and Curley (43) recognized the need for the development of a better understanding and definition of redundancy in various types of bridges. They defined the term R as the redundant factor for a bridge, which is the reserve strength between component(s) damage and sys- tem collapse. The redundancy factor was later used in other studies [e.g., Frangopol and Curley (44), Frangopol and Nakib (45), Frangopol and Yoshida (46)] to investigate redundancy of systems. Ghosn and Moses (34) included both a reliability approach and a direct system factor approach to evaluate the degree of redundancy of an existing bridge or when designing a new bridge. In the reliability approach, relative reliability was calculated and a level of redundancy is satisfied if obtained values of the relative reliabilities are greater than or equal to specified values. In the direct system redundancy approach, adequate load factor ratios (system reserve ratios) are required FIGURE 16 View of cracked girder in two-girder span of Lafayette Street Bridge in St. Paul, Minnesota, as an example of a bridge that is sufficiently redundant to avoid collapse despite a fracture of the tension flange and the web of one girder. FIGURE 17 View of cracked girder in two-girder span of I-79 Bridge at Neville Island in Pittsburgh as an example of a two-girder bridge that is sufficiently redundant to avoid collapse despite a fracture of the tension flange and the web of one girder. FIGURE 18 Example of train derailment on a fracture-critical truss bridge that severed several members but did not collapse. (Courtesy: Robert Sweeney.)

19 to satisfy a minimum level of redundancy. A bridge is then considered adequate if system reserve ratios are greater than or equal to specified values. A drawback of the reliability approach is that it requires measures of statistical variation that are often not available. In these examples, estimates are made and the results are still insightful, but a great deal of additional data would be required to use this approach as a practical tool. It is important to note that it is the reliability of the system that is important. As discussed previously, redundancy affects this reliability but not as much as might be assumed. A two- girder bridge designed for HS-25 for example might have greater reliability than a multigirder bridge designed for HS-20 (34). Therefore, one should not place too much empha- sis on redundancy and lose sight of the important goal, sys- tem reliability. Numerical Simulations of the Residual Capacity of Fractured Bridges Finite-element analysis is increasingly being used to simu- late the after-fracture behavior and residual capacity of FCBs. These analyses provided insight about the secondary load paths in FCB systems after an FCM is severed or otherwise removed from the model. In some cases they are used to get a waiver from FHWA on FCM design requirements for a new bridge. However, this type of analysis and associated waiver of the FCM provisions is presently being done on a case-by-case basis, and the analysis requirements, loads, and failure criteria are not always clear. In other cases, they are used to evaluate the residual capacity of existing FCBs. Section 6.6.2 of the AASHTO LRFD Bridge Design Spec- ifications discusses the fracture limit state. The commentary of this section states that: The criteria for refined analysis used to demonstrate that part of a structure is not fracture critical, has not yet been codified. Therefore, the loading cases to be studied, location of potential cracks, degree to which the dynamic effects associated with a fracture are included in the analysis, and fineness of the models and choice of element type should all be agreed upon by the owner and the engineer. The ability of a particular software product to adequately capture the complexity of the problem should also be considered and the choice of software should be mutually agreed upon by the owner and the engineer. Relief from full fac- tored loads associated with the Strength I Load Combinations of Table 3.4.1-1 should be considered as should the number of loaded design lanes versus the number of striped traffic lanes (1). Heins and Hou (47) used an analytical two-girder and three-girder space frame model to study the effect of bracing members in bridge structures on the load distribution of two- girder and multigirder systems after the development of a crack in one of the girders. Heins and Hou found that when one crack develops and no bracing is used, the deformation increases by 40% for the two-girder system and 10% for the three-girder system. However, if bracings are considered, the deformation increases by 10% for the two-girder system, whereas almost no increase in the deformation is noticed in the three-girder system. Heins and Kato (48) also studied the effect of bracing on load distribution in a two-girder bridge system when one of the girders is damaged and found that the deformation of the cracked girder was substantially reduced when bracing was used. Ghosn and Moses (34) developed recommendations for the residual capacity of fractured bridges to demonstrate sufficient redundancy. Load factors LF1 and LFd should be calculated using a three-dimensional finite-element analy- sis. LF1 is the multiple of side-by-side HS-20 trucks that the structure can carry in addition to unfactored dead loads (using elastic analysis) before the first member reaches the resistance predicted by the design specifications. LF1 is typ- ically on the order of 3.8, depending on the ratio of live load to dead load LFd is the residual capacity of the damaged structure and is calculated by performing a nonlinear analysis of the dam- aged structure (with the FCM removed) under the effect of the unfactored dead load and incrementing the multiple of side-by-side HS-20 truck loads until the system collapses. Redundancy is considered adequate if the ratio of LFd to LF1 is greater than 0.5. This means that the damaged structure should be able to support approximately 1.9 times the side- by-side HS-20 loading. (Figure 19 shows a schematic of the loading of a damaged girder in terms of multiples of side-by- side HS-20 trucks.) Note that this requires a greater residual LFd if the bridge is originally over-designed, which does not seem logical. NCHRP Report 406 indicated that bridges that do not meet the required load factor ratios could still provide a high level of system safety (34). FIGURE 19 Schematic of multiple HS-20 loads on damaged superstructure. (Courtesy: HNTB.)

HNTB, as part of Milwaukee Transportation Partners, used these recommendations in a redundancy analysis for twin curved box girders designed for the Marquette Interchange in Wisconsin to demonstrate to FHWA that waiving the fracture-critical inspection requirements was justified (49). The boxes were going to be fabricated as FCMs anyway. (Recall previously in this chapter that the fabrication part of the additional cost of FCMs is relatively small compared with the in-service inspection part.) A three-dimensional shell element model was used for elastic incremental analyses, manually updating the mesh to account for nonlinear effects. The reserve capacity of the undamaged bridge was found to be 4.9 times HS-25 side-by- side trucks for the pier section of the outside box and 3.4 times HS-25 side-by-side trucks for the midspan section of the outside box. The residual capacity of the damaged midspan section was found to be 3.35 times HS-25 side-by-side trucks, above the 0.5 times the original reserve capacity recommended by Ghosn and Moses (34). A dynamic amplification factor of 1.8 was conservatively estimated using a single-degree-of-freedom impact model and assuming 5% of critical damping. Based on this, it was deter- mined that the structure might have to withstand dynamic stresses that are equivalent to 2.68 times HS-25 side-by- side trucks in addition to the static dead load effect. (The 2.68 level included the anticipated dynamic part of the dead load and both the static and dynamic part of one set of HS-25 trucks as the live load.) Because the residual capacity was greater than 2.68, the bridge is considered safe for the tem- porary dynamic loading. Lai (50) developed a three-dimensional finite-element model and used a static incremental study to determine the degree of redundancy of a tied arch bridge. He found that after the fracture of the one of the ties, the structure was capable of carrying its own weight plus 1.3 times HS-20 truck load- ing without catastrophic collapse. Michael Baker Jr., performed an analysis for the proposed Blennerhassett Tied Arch Bridge in West Virginia (51). The tie girder section is a bolted built-up box section as shown in Figure 10b. Because the section is a bolted built-up section, it has internal redundancy as discussed previously in this chapter. In this case, a fatigue crack, fracture, corrosion fail- ure, ductile rupture, or other failure of any one of the four plates cannot propagate directly into any of the other three plates. The designer justified using the inelastic capacity of the residual cross section because AASHTO LRFD Specifica- tions Articles 4.5.2.1 and 4.5.2.3 permit inelastic response for extreme events, and the fracture of one of the plates in the cross section was considered an extreme event (51). The analysis of the Blennerhassett Bridge did not include the effect of plastic redistribution of forces, which, if included, indicate greater capacity. 20 The tie girder was modeled using elastic beam elements. Elastic analysis is always conservative, even when used with inelastic cross-section capacity, provided the sections are sufficiently ductile. The analysis involved a 14-ft (4.2-m)- long segment of the tie girder representing a fractured resid- ual section. The centroid of the C-shaped residual section is not collinear with the centroid of the uncracked tie girder, creating some additional bending moment. The capacity of the residual section is then compared with the elastic moments from this model with the full-factored AASHTO LRFD Specifications HL-93 load and impact [1.25 DC + 1.50 DW + 1.30 (LL+I)]. Although no dynamic amplifica- tion effects were included, the Strength I factored loading is very conservative and would amount to a large multiple of HS-20 trucks. The capacity of the cross section is obtained from a model that relates moment to curvature and strain. Nominal cross- section strains of less than 1%, approximately six times the yield strain of the HPS 70W steel, were considered accept- able. This level of allowable nominal strain allows a margin for localized concentrated strains near holes and other dis- continuities, and was supported by nonlinear finite-element analysis and testing of tension and flexural members made from similar HPS 70W at the University of Minnesota (52,53). The residual cross section; that is, a cross section with either a web or a flange missing, was shown to be able to continue to carry the load after such an event. Therefore, the tie girder was shown to meet the definition of redundant. This same type of analysis is being used to evaluate older structures, as well to better direct resources for maintenance and replacement; for example, if a bridge is not really fracture- critical then it may not be necessary to replace it as soon. For example, as part of a study being conducted on the I-35W truss bridge in Minneapolis, Minnesota, complex three-dimensional finite-element models of the bridge are being developed. Fracture-critical members are removed from the model and the resulting redistribution of loads accurately is being studied. Adequacy of connections, individual components, and over- all stability are being assessed. PRESENT CLASSIFICATION OF STEEL BRIDGE SUPERSTRUCTURES AS FRACTURE-CRITICAL Common assumptions that fractures in certain superstructure types will lead to collapse may be too simplistic, as shown by the bridges traditionally classified as FCBs that did not col- lapse, as discussed earlier in this chapter. Table 1 is a chart assembled by the California DOT (Cal- trans) showing the varying definitions of different superstruc- ture types as “fracture-critical” in four related documents. There appears to be substantial disagreement, as was also found in the survey conducted as part of this project (see chap- ter three). Note that only the book (in the left-hand column) considers two-box-girder systems to be FCBs. In the case

21 Bridge Inspection and Structural Analysis (27) Manual for Condition Evaluation and Load and Resistance Factor Rating of Highway Bridges (3) Inspection of Fracture Critical Bridge Members, Report FHWA-IP-86-26 (28) Bridge Inspectorís Reference Manual (29) Two-girder system (simple and continuous span) • Suspended span with pin and hanger system • Suspended span • Welded plate girder • Riveted or bolted plate girder One- or two-girder systems, including single boxes with welding Suspended spans with two girders Two-girder systems (single span and end span of continuous span units) • With fix hanger suspended span • With suspended span • Welded plate girders • Riveted or bolted plate girders • Suspended span with two girders • Simple span two-girder bridge with welded partial length cover plates on the bottom flange • Continuous span two-girder system with cantilever and suspension link arrangement and welded partial length cover plates • Simple span two-girder system with lateral bracing connected to horizontal gusset plates that are attached to webs Truss system (simple and continuous span) • Eyebar truss • Welded truss • Riveted truss • Three deck truss with pin and hanger assembly Two-truss systems Truss system (simple and continuous spans) • Eyebar truss • Welded truss • Truss with suspended span • Riveted truss • Simple span truss with two eyebars or single member between panel points Suspension bridge • Eyebar chain • Cable Suspension systems with two eyebar components Suspension bridge • Eyebar chain • Cable • Bar chain suspension bridge with two eyebars per panel TABLE 1 VARIOUS DEFINITIONS OF COMMON FRACTURE-CRITICAL BRIDGE SUPERSTRUCTURES (continues)

22 • Welded tie box girder • Riveted tie box girder Welded tie arches • Two welded tie box girder • Two riveted tie box girder Steel pier cap • Welded box or plate girder • Riveted box or plate girder • Two column steel bent Steel pier caps and cross girders Steel pier cap • Welded box or plate girder • Riveted plate girder • Single welded I-girder or box girder pier cap with bridge girders and stringers attached by welding Longitudinal box beam • Single welded box • Single riveted box Single boxes with welding Longitudinal box beam • Single welded box • Single riveted box • Simple span single welded box girders with details such as termination of longitudinal stiffeners or gusset plate Anchor for cable stayed bridge Pin and hanger connections when used in suspended span configuration in nonredundant systems Pin and hanger connections on two- or three- girder systems Two or less box girders Three or more box girders if spacing is large (should be determined by structural engineer) Courtesy: Tom Harrington, Caltrans. Tie arch Tie Arch Welded tie arches with box shape tie girder TABLE 1 (Continued)

23 of the AASHTO Manual for Condition Evaluation and Load and Resistance Factor Rating (LRFR) of Highway Bridges (3) [identical to the predecessor Manual for Condition Evalua- tion of Bridges (54)] and the Bridge Inspectors Reference Manual (29), only welded tie members for arches or single box girders are considered FCMs, whereas riveted or bolted built-up tie members or single box girders, such as those shown in Figure 10, would not be FCMs. This would appear to be giving credit to the internal redundancy of the bolted or riveted built-up members (but these manuals note elsewhere that internal redundancy should be neglected). In Table 1, not all two-I-girder bridges are considered FCBs in some of the documents. The LRFR Manual and the reference manual include all two-I-girder bridges as FCBs. However, in Inspection of Fracture Critical Bridge Members (28), continuous two-girder bridges are not considered frac- ture-critical except for the end span. This FHWA report gives credit to the structural redundancy of the continuous spans. Unfortunately, the AASHTO LRFD Bridge Design Speci- fications are not clear about what types of superstructures are to be classified as FCBs. The fatigue and fracture limit state requirements are specified in Article 6.10.5. Special provi- sions for box sections are included in Section 6.11.5, which also states that: For single box sections, box flanges in tension shall be consid- ered fracture-critical, unless analysis shows that the section can support the full dead and appropriate live load after sustaining a complete fracture of the flange at any point. The commentary of this section states that: There may be exceptions where box flanges of single-box sections subject to tension need not be considered fracture-critical. For example, continuously braced top flanges in regions of nega- tive flexure where there is adequate deck reinforcing to act as a top flange in such cases, adequate shear connection must also be provided. The section was recently amended to include the following: Unless adequate strength and stability of a damaged structure can be verified by refined analysis, in cross sections comprised of two box sections, only the bottom flanges in the positive moment region should be designated fracture-critical. Where cross sections contain more than two box girder sections, all of the tension flanges should be considered non-fracture-critical. Therefore, the redundancy of two box or tub girders is some- times even in question. Twin tub girders would be expected to perform even better than twin I-girders owing to the torsional capacity of the intact tub girder and the alternate load paths available within each tub girder. Unfortunately, this built-in redundancy shown by these structures is not explicitly rec- ognized in the LRFD Specifications. As will be discussed in chapter three, the results of the survey revealed that different agencies classify tub girders differently when determining if the bridge is fracture-critical or not.

Next: Chapter Three - Results of Survey »
Inspection and Management of Bridges with Fracture-Critical Details Get This Book
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TRB’s National Cooperative Highway Research Program (NCHRP) Synthesis 354: Inspection and Management of Bridges with Fracture-Critical Details explores the inspection and maintenance of bridges with fracture-critical members (FCMs), as defined in the American Association of State Highway and Transportation Officials’ Load and Resistance Factor Design (LRFD) Bridge Design Specifications. The report identifies gaps in literature related to the subject; determines practices and problems with how bridge owners define, identify, document, inspect, and manage bridges with fracture-critical details; and identifies specific research needs. Among the areas examined in the report are inspection frequencies and procedures; methods for calculating remaining fatigue life; qualification, availability, and training of inspectors; cost of inspection programs; instances where inspection programs prevented failures; retrofit techniques; fabrication methods and inspections; and experience with fracture-critical members fractures and problems details.

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