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39 APPENDIX A Background Discussion on Fatigue, Fracture, Nondestructive Evaluation, and Repair and Retrofit OVERVIEW OF FATIGUE the nominal stress range in the connecting members rather than the local "concentrated" stress at the detail. The nominal stress Fatigue is considered a serviceability limit state for bridges. is usually obtained from standard design equations for bending This is because fatigue cracks do not typically compromise and axial stress and does not include the effect of stress con- structural integrity and are more of a maintenance issue. centrations of welds and attachments. AASHTO (A2) has seven However, as was recognized by the Task Committee on SN curves corresponding to seven categories of weld details Redundancy of Flexural Systems (A1), fatigue is the most (A through E'), as shown in Figure A1. common cause of reported damage in steel bridges. The fatigue design procedure is based on associating the The fatigue design and assessment procedures outlined in weld detail under consideration with a specific category. The this appendix are included in the AASHTO specifications for effects of the welds and other stress concentrations, includ- bridges (A2). As a result, steel bridges that have been built in ing the typical defects and residual stresses, are reflected in the last two decades have not and should not have any signifi- the ordinate of the SN curves for the various detail cate- cant problems with fatigue and fracture (A3). However, bridges gories. Consequently, the variability of fatigue life data at a designed before the modern specifications will continue to be particular stress range is typically about a factor of 10. susceptible to the development of fatigue cracks and to fracture. The AASHTO SN curves in Figure A1 are also used Detailing rules are perhaps the most important part of the throughout North America for a variety of other welded fatigue and fracture design and assessment procedures. structures, including the American Institute for Steel Con- struction Manual of Steel Construction (A4), the American These rules are intended to avoid notches and other stress Railway Engineering and Maintenance-of-Way Association concentrations, as well as the use of details known to be very Manual for Railway Engineering (A5), the American Weld- fatigue sensitive. They also often result in details that have ing Society (AWS D1.1) Structural Welding Code (A6), and improved resistance against brittle fracture as well as fatigue. the Canadian Standards Association (CSA S16-2001) Limit Modern steel bridges are also detailed in a way that appears States Design of Steel Structures (A7). much cleaner than those built before the 1970s. There are fewer connections and attachments in modern bridges and The SN curves can be represented by the following power the connections use more fatigue-resistant details, such as law relationship: high-strength bolted joints. A N= (1) In bridges there are usually a large number of cycles of sig- S3 nificant live load and fatigue will almost always precede frac- ture. Therefore, controlling fatigue is practically more impor- where N is the number of stress cycles, S is the nominal stress tant than controlling fracture. Usually, the only measures taken range, and A is a constant particular to the detail category, as in design that are primarily intended to ensure fracture resis- given in Table A1. tance are those to specify materials with minimum specified toughness values [such as a Charpy V-notch (CVN) test re- In the nominal stress range approach, each detail category quirement]. As explained in chapter three, toughness is speci- also has a constant-amplitude fatigue limit (CAFL), which is fied so that the structure is resistant to brittle fracture despite given in Table A1. The CAFL is the stress range below which manufacturing defects, fatigue cracks, and/or unanticipated no fatigue cracks occurred in tests conducted with constant- loading. However, these material specifications are less impor- amplitude loading. tant for bridges than the SN curves and detailing rules. The approach to designing and assessing bridges for fatigue is empirical and is based on tests of full-scale mem- Nominal Stress SN Curves bers with welded or bolted details. Such tests indicate that: The established method for fatigue design and assessment of The strength and type of steel have only a negligible steel bridges in the United States is the nominal stress approach. effect on the fatigue resistance expected for a particular The nominal stress approach is based on SN curves, where S detail (A8A10). is the nominal stress range and N is the number of cycles until The welding process also does not typically have an the appearance of a visible crack. Details are designed based on effect on the fatigue resistance (A11,A12).

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40 AASHTO Curves 1000 100 Stress Range, MPa Stress Range, ksi 100 10 10 104 105 106 107 108 Number of Cycles FIGURE A1 Lower-bound SN curves for seven primary fatigue categories from AASHTO specifications. Dotted lines are the constant-amplitude fatigue limits and indicate detail category. The primary effect of constant-amplitude fatigue loading ture-critical and non-fracture-critical bridges (FCBs and non- can be accounted for in the live-load stress range (A8 FCBs). However, the concern is greater with respect to frac- A10); that is, the mean stress is not significant. (The rea- ture-critical structures, as indicated by several responses to son that the dead load has little effect is that, locally, there the survey and as discussed in chapter three. are very high residual stresses.) The full-scale fatigue experiments upon which the fatigue It is worth noting that when information about a specific rules are based were carried out in moist air and therefore crack is available, a fracture mechanics crack growth rate reflect some degree of environmental effect or corrosion analysis should be used to calculate remaining life (A13 fatigue. Hence, the lower-bound SN curves in Figure A1 can A16). However, in the design stage, without specific initial be used for the design of details with a mildly corrosive envi- crack size data, the fracture mechanics approach is not any ronment (such as the environment for bridges, even if salt or more accurate than the SN curve approach (A17 ). Therefore, other corrosive chemicals are used for deicing) or provided there will be no further discussion of the fracture mechanics with suitable corrosion protection (galvanizing, other coat- crack growth analysis. ing, or cathodic protection). Some design codes for offshore structures have reduced fatigue life by approximately a fac- Effect of Corrosion on Fatigue Resistance tor of two when details are exposed to seawater (A18,A19). Many bridge owners are concerned about the influence of cor- At the relatively high stress ranges at which most acceler- rosion on the fatigue and fracture performance of both frac- ated tests are conducted, the effect of seawater is clearly detri- mental. However, there is evidence that the effect of corrosion in seawater is not so severe for long-life, variable-amplitude TABLE A1 fatigue of welded details. Full-scale fatigue experiments in sea- PARAMETERS FOR SN CURVES water at realistic service stress ranges do not show significantly AASHTO Coefficient A CAFL lower fatigue lives, provided that corrosion is not so severe that Category (MPa3) (MPa) it causes pitting or significant section loss (A20). At relatively A 81.9 x 1011 165 low stress ranges near the CAFL typical of service loading, it B 39.3 x 1011 110 appears that the build-up of corrosion product in the crack may B 20.0 x 1011 83 C 14.4 x 1011 69 actually increase crack closure and retard crack growth, at least D 7.21 x 1011 48 enough to offset the increase that would otherwise occur owing E 3.61 x 1011 31 to the environmental effect. The fatigue lives seem to be more E' 1.28 x 1011 18 significantly affected by the stress concentration at the toe of CAFL = constant-amplitude fatigue limit. welds than by the corrosive environment.

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41 Severely corroded members may be evaluated as Category is to provide procedures for calculating the remaining fatigue E details (A21), regardless of their original category (unless life of existing steel bridges using concepts of probabilistic of course they were Category E or worse to start with). How- limit states. The fatigue evaluation procedures in the guide ever, pitting or significant section loss from severe corrosion were adopted from and are identical to the proposed proce- can also lower that fatigue strength (A21,A22). dures presented in NCHRP Report 299: Fatigue Evaluation Procedures for Steel Bridges (A26). The effects of repeated Loading for Fatigue Design loading on the fatigue life of an existing bridge are defined in terms of the remaining life of the structure. This means that It is important to note that the load producing fatigue cracking the effect of exceeding the allowable fatigue stress is a reduc- is comprised of the entire load spectrum crossing the bridge. tion of the remaining fatigue life of the structure rather than The load that produces a fracture, on the other hand, is typi- immediate failure. cally the largest load from the load spectrum. Bridges experi- ence what is known as long-life, variable-amplitude loading There are two levels for which the remaining fatigue life [i.e., very large numbers of random amplitude cycles greater can be calculated, the remaining safe life and the remaining than the number of cycles associated with the CAFL (A23)]. mean life. The remaining safe life provides a much higher level of safety. The safe life has an exceedence probability of If the percentage of stress ranges exceeding the CAFL is 97.7% for redundant members (the same probability inherent greater than approximately 0.01%, the history of N variable in using the basic design SN curves). The mean remaining stress ranges can be converted to N cycles of an effective stress life, however, is the best estimate of the actual remaining range that can then be used just like a constant-amplitude stress fatigue life of the detail under consideration. The mean life range in SN curve analysis. Typically, Miner's rule (A24) is has an exceedence probability of 50%. used to calculate an effective stress range from a histogram of variable stress ranges. Theoretically, this effective constant- The safe life is used for design and for a first screening amplitude stress range results in approximately the same analysis. The impact of reaching a safe life equal to zero fatigue damage for a given number of cycles as the same num- should be relatively minimal; it may mean having to repair ber of cycles of the variable-amplitude service history. If the just a small percentage of the details on a bridge. In this case, stress ranges are counted in discrete "bins," as in a histogram, the calculated mean life may be used to determine when the effective stress range, SRe (A23), can be calculated as: the cracking will be so pervasive that half the details will be cracking, requiring extensive retrofitting or possibly ( ) 13 replacement of the bridge. If the remaining life that is esti- SRe = ( S ) i i 3 ri (2) mated is not satisfactory, there are four options: (1) recalcu- late the fatigue life more accurately (possibly using load test- where i = number of stress cycles with stress range in the ing), (2) restrict truck traffic on the bridge, (3) repair or modify bin with average value Sri divided by the total number of the detail, and/or (4) perform more frequent inspections of stress cycles (N). the detail. In the AASHTO specifications (A2), the stress range from The procedures in NCHRP Report 299 (A26) are the best the fatigue design truck (i.e., the HS15) represents the effec- available procedures for predicting the number of years tive stress range. No additional safety factor is used for before substantial fatigue cracking may occur at a detail. It is Miner's rule, because it is relatively accurate for truck loading difficult to accurately predict all of the variables that the on bridges. For large numbers of cycles, the AASHTO speci- bridge will experience over its remaining life, such as past, fication has another check that involves comparing the stress present, and future truck volumes and stress ranges, which range from the fatigue design truck with one-half of the CAFL. can also vary with changing truck weights over time. This check is actually intended to compare the CAFL with a stress range that is twice that produced by the fatigue design Although calculations are by far the most common and truck (i.e., dividing the resistance by two is the same as multi- favored methods of estimating remaining fatigue life in a plying the load by two). Although somewhat confusing in bridge, they are also the least accurate. Field instrumenta- application, the intent is to ensure that almost all of the stress tion and monitoring has consistently demonstrated that mea- ranges should be below the CAFL, but that occasionally the sured in-service stress range histograms result in the most stress range can exceed the CAFL with no significant effect. accurate estimates of remaining life. Analysis methods rely on approximate load models and simplified structural analy- Fatigue Life Prediction Methodology sis models, both of which are typically conservative. Mea- for Existing Bridges surements made in the field reflect actual site conditions and traffic patterns when recorded over a sufficient period of The 1990 AASHTO Guide Specifications for the Fatigue time. Because the stress range is directly measured at the Evaluation of Existing Steel Bridges (A25) can be used for the detail in question, there is no error introduced through ana- fatigue evaluation of steel bridges. The purpose of the guide lytical simplifications.

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42 Because in nearly all cases field measurements indicate Fracture Assessment Procedures that in-service stresses are less than predicted, costly retrofits can be eliminated or reduced in number. The resulting sav- A fracture assessment is made using fracture mechanics prin- ings will almost always far exceed the cost of the additional ciples (A1315,A32,A33). In the absence of other established efforts associated with the field instrumentation. Despite this procedures, an often used reference is the 1999 British Stan- potential for savings, most agencies do not use this approach, dard, BS 7910, "Guide on Methods for Assessing the as will be discussed in the review of the survey results. Acceptability of Flaws in Metallic Structures" (A32). There is presently no comparable U.S. standard; however, BS 7910 is used by some U.S. industries (primarily the oil and gas OVERVIEW OF FRACTURE industries). Fracture may be defined as rupture in tension or rapid exten- A fracture assessment requires some knowledge of the sion of a crack, leading to gross deformation, loss of function fracture toughness of the steel and/or weld metal, which may or serviceability, or complete separation of the component be estimated from the Charpy energy or CVN (A15,A33). (A4,A14,A27). Because the scope of this synthesis is limited Unlike fatigue, the susceptibility to fracture is strongly to practical information, there are many interesting aspects of dependent on the type of material and even the particular heat fracture that are not discussed. However, there are several of the steel or lot of weld metal (A14,A15,A34,A35). The good texts that can serve as a starting point for more in-depth CVN may be obtained from mill reports for the steel and studies (A14,A15,A27). from the certifications for the weld metal, if good records It is important to rapidly assess the potential for fracture exist. Lacking such specific information, for bridges built whenever there is a crack in any tension element (tension since 1974 when minimum Charpy requirements were first member or tension flange of flexural member). However, the included in the ASTM A709 specification for bridge steels, occurrence of fatigue cracks does not necessarily mean that the CVN for the steel plate and weld metal may be assumed the structure is in danger (A13,A27). In some redundant to be at least as large as the minimum specified values. For structures, a fatigue crack may stop propagating with no older structures, or for cases when the assumed minimum intervention at all as a result of redistribution of stresses values are not sufficient, it may be necessary to drill a core (A28). Usually, however, a fatigue crack will propagate and and get a sample of the steel, then make and test some Charpy eventually cause a fracture if not repaired in a timely manner specimens. A minimum of three specimens should be tested (A13,A15,A27). The development of a fatigue crack in a frac- from each plate or structural shape. The fracture assessment ture-critical member (FCM) should, in most cases, warrant will not be discussed further because it is documented else- closure of one or more lanes, posting the bridge for cars only, where and is not the emphasis of this report. or result in complete closure of the bridge until repairs can be made. Fracture is the rupture in tension or rapid extension Causes for Fracture Other Than Fatigue of a crack leading to gross deformation, loss of function or serviceability, or complete separation of the component. As mentioned previously, fracture in bridges is almost always a result of fatigue cracking. However, if a detail and material Details that have good fatigue resistance, Category C and are particularly susceptible to fracture, failure will usually better, are usually also optimized for resistance to fracture. occur as the loads are applied for the first time during or soon Detailing rules to avoid fracture are very similar to the after construction. An example is the fracture that occurred common sense rules to avoid fatigue. For example, inter- when the Caltrans Workers Memorial Bridge (previously secting welds should always be avoided owing to the proba- known as the Bryte Bend Bridge) was under construction in bility of defects and excessive constraint. Intersecting welds, 1970. The fracture, shown in Figure A2, was attributed to use or even welds of too close proximity, have caused brittle of low-toughness A514 steel (A36) at a point where a trans- fractures [e.g., the Hoan Bridge in Wisconsin (A29) and the verse bracing member was welded along the edge of the pri- SR-422 bridge in Pennsylvania (A30,A31)]. Weld backing mary tension flange of a tub girder, creating a stress concen- bars must usually be removed to achieve the needed resis- tration at the reentrant corner. A514 steel, marketed under the tance to both fatigue and fracture. Stress concentrations such name of T1 steel, is quenched and tempered with a minimum as reentrant corners should be avoided and instead transition specified yield stress of 690 MPa. radii that are ground smooth and flush should be provided. This commonly occurs at copes in floorbeams at connections It is also possible that fracture can occur in service directly designed for vertical shear only. without apparent fatigue crack growth. For example, at poor details that are highly constrained, such as the intersection point The ends of butt welds are always a potential location of of two or three welds, fracture may occur in service directly defects. It is important to use run-out tabs and to later grind from small crack-like weld discontinuities, such as the fractures the ends of the weld flush or to a radius. Fillet weld termina- that originated at shelf plate details in the Hoan Bridge in Mil- tions should not be ground, for this will expose a very thin waukee in December 2000 (A29) (Figure A3) or the SR-422 ligament near the weld root that will tear easily. failure (A30,A31). Repair and retrofit of these shelf-plate details

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43 influence of tensile stress (A14,A15). As hydrogen is liber- ated in the process, this failure mechanism may also involve hydrogen cracking. SCC can be distinguished from fatigue cracking by examining the fracture surface under a light microscope. SCC has occurred in prestressing cables (A37) and in A490 bolts that were out of specification (A38). Specification of Charpy V-Notch for Fracture Resistance Fracture behavior depends strongly on the type and strength level of the steel or filler metal. The fracture resistance of each type of steel or weld metal varies significantly from heat FIGURE A2 Brittle fracture of flange of tub girder of Bryte Bend to heat, or from lot to lot. Although fracture toughness can be Bridge in Sacramento, California, while under construction. measured directly in fracture mechanics tests (A14A16), the usual practice is to characterize the toughness of steel in terms of the impact energy absorbed by a CVN specimen are discussed in special directives from FHWA and are not dis- (A15). Because the Charpy test is relatively easy to do, it will cussed further in this report. likely continue to be the measure of toughness used in steel specifications. Subcritical crack propagation in bridge elements may also occur by stress corrosion cracking (SCC), although this is a Because it is not directly related to the fracture toughness, concern only for very-high-strength steels; that is, steel with CVN energy is often referred to as notch toughness. The yield strength much greater than 100 ksi. SCC involves elec- notch toughness is still very useful, however, because it can trochemical dissolution of metal along active sites under the often be correlated to the fracture toughness and then used in a fracture mechanics assessment (A15,A32,A33). Figure A4 shows a plot of the CVN energy of A709 Grade 50 (350 MPa yield strength) structural steel at varying temperatures. These results are typical for ordinary hot-rolled structural steel. The fracture limit state includes phenomena ranging from the brittle fracture of low-toughness materials at service load levels to ductile tensile rupture of a component. The transi- tion between these phenomena depends on temperature, as reflected by the variation of CVN with temperature as shown in Figure A4. The transition is a result of changes in the underlying microstructural fracture mode. Brittle fracture on the so-called lower shelf in Figure A4 is associated with cleavage of individual grains on select crystallographic planes. Brittle fracture may be analyzed with linear-elastic fracture mechanics theory because the plastic zone at the crack tip is very small. At the high end of the temperature range, the so-called upper shelf, ductile frac- ture is associated with the initiation, growth, and coalescence of microstructural voids, a process requiring much energy. The net section of plates or shapes fully yields and then rup- tures with large slanted shear lips on the fracture surface. Transition-range fracture occurs at temperatures between the lower and upper shelves and is associated with a mixture of cleavage and shear fracture. Large variability in toughness at constant temperature and large changes of temperature are typical of transition-range fractures. FIGURE A3 Fractured girder of the Hoan Bridge in Milwaukee (top) and view of critical shelf plate detail featuring intersecting AASHTO specifications for bridge steel and weld filler welds. metal require minimum CVN values at specific temperatures

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44 TEST TEMPERATURE, F -150 -100 -50 0 50 100 150 200 250 90 A588 19mm Plate Charpy Data 60 70 CVN, Joules CVN, ft-lb 40 50 30 20 10 -75 -50 -25 0 25 50 75 100 TEST TEMPERATURE, C FIGURE A4 Charpy energy transition curve for A709 Grade 50 (350 MPa yield strength) structural steel. (A25). As shown in Figure A4, the typical lower-shelf CVN is is well above the lower shelf; however, there may be a greater about 10 J. Therefore, when a minimum CVN of 20 J or more premium to be paid with diminishing increases in certainty. is specified at some temperature, the most important result of such a specification is that the lower shelf of the Charpy Table A2 lists, as an example, values of impact energy curve will start at a temperature lower than the specified tem- presently required for base metal in FCMs for the three tem- perature. This indicates that the lower shelf of a structure perature zones into which the United States is divided. [For loaded statically or at intermediate strain rates such as traffic nonfracture-critical members (non-FCMs), typically only 20 loading on a bridge is even lower, a phenomena known as the J is required at the same temperatures.] The complete table is temperature shift (A15). Because of the temperature shift, the found in the LRFD Specifications (A2). For example, 50-mm temperature at which the CVN requirement is specified may thick flange plates of grade 345W steel for an FCB to be built be greater than the lowest anticipated service temperature. at a site where the lowest anticipated service temperature (LAST) is -15C must have the minimum impact energy for If the material is not on the lower shelf at service tempera- Zone 1, which is 34 J at 21C. ture, brittle fracture will not occur as long as large cracks do not develop. It essentially does not matter what the specified History of Development of the Fracture CVN value is as long as it is at least 20 J. Usually, an average Critical Plan from three tests of 34 J (25 ft-lbs) or 27 J (20 ft-lbs) is speci- fied at a particular temperature. The greater the value of the The fracture toughness requirements are based on a correla- average CVN requirement, the more certain that the material tion between the fracture toughness in terms of the stress- TABLE A2 MINIMUM CHARPY IMPACT TEST REQUIREMENTS FOR FRACTURE-CRITICAL MEMBERS Plate Thickness (mm) Zone 1 Zone 2 Zone 3 ASTM A709 and LAST = -18C LAST = -34C LAST = -51C Steel Grade Joining Method a (Joules at C) (Joules at C) (Joules at C) 250F Up to 100 M & W 34 at 21 34 at 4 34 at -12 Up to 50 M & W 34 at 21 34 at 4 34 at -12 345F, 345WF Over 50 to 100 M 34 at 21 34 at 4 34 at -12 Over 50 to 100 W 41 at 21 41 at 4 41 at -12 HPS-485WF Up to 100 M & W 48 at -23 48 at -23 48 at -23 Up to 65 M & W 48 at -1 48 at -18 48 at -34 Over 65 to 100 M 48 at -1 48 at -18 48 at -34 690F, 690WF Over 65 to 100 W 68 at -1 68 at -18 Not permitted a M = mechanically fastened; W = welded.

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45 intensity factor and the CVN, including a "temperature ferent than mild steel and should be avoided for FCMs. Frac- shift." The temperature shift accounts for the differences in tures that occurred during construction in the box girders of the strain rate between the Charpy impact test and the traffic Bryte Bend Bridge in California also occurred in A514 steel. loads. For example, the time it takes to reach the yield strain Consequently, higher CVN requirements are implemented for in bridges from an overloaded truck is about one second ver- the higher strength steel, as shown in Table A2. sus one millisecond in the Charpy test (A15). Several inde- pendent studies verified the temperature shift (A39A41). Again a compromise was reached, which resulted in the slightly increased CVN for FCMs at the same temperature For example, several girders were tested at the U.S. Steel shift. At the same time, more stringent fabrication rules were Research Laboratory that contained fatigue cracks initiated at applied for FCMs to control the possibility of initial defects. the ends of cover plate details (A41). A more extensive study Most importantly, the stress range for fatigue design was was done at Lehigh University that examined an increased arbitrarily dropped by about one category, which resulted in number of specimens fabricated from grade 250, 345, and 690 a decrease in the allowable stress range of approximately steels (A39). In both studies, the fatigue cracks were grown at 25% in most cases. A 25% decrease in the stress range should temperatures and loading rates that closely resemble those essentially double the fatigue life. It is important to realize encountered in bridges, and the cracked girders were sub- that this decrease in the allowable stress range for FCMs in jected to periodic overloads until fracture occurred. the Standard Specifications has been dropped in the most recent LRFD Specifications (A2). The temperature shift was the subject of some disagreement in the development of notch toughness requirements in the The Bridge Welding Code D1.5 calls for higher impact early 1970s. The prevailing opinion in the debate was that energies in weld metal than in base metal, a reasonable ordinary hot-rolled mild structural steels had sufficient notch requirement given that the welds create stress raisers and toughness. The specified CVN values are the typical minimum high-tensile residual stresses. Given that the cost of filler values that could be consistently achieved. Having the speci- metal is relatively small in comparison with the overall cost fication was very important to screen out unusually brittle of materials, it is well worth specifying high-toughness filler materials. metal to mitigate the effects of high-tensile stresses induced by welding. For FCMs, 34 J is required at -29C for weld Roberts and Krishna (A39) and Roberts et al. (A40), metal, regardless of the temperature zone. For non-FCMs, among others, argued that the fracture-critical specification weld metal is required that can give 27 J at -29C for Zone should require a level of dynamic toughness sufficient to 3, and at -18C for Zones 1 and 2. arrest pop-in-type fractures occurring from local brittle zones. Hartbower (A42) proposed an alternate fracture control plan The somewhat higher toughness requirements shown in that called for the CVN test to be done at the same tempera- Table A1 for thick plates that are to be welded is the result of tures as the LAST for the bridge service site; that is, without the higher degree of constraint and planestrain behavior at exploiting the benefit of the temperature shift. This would temperatures higher than would be the case for thinner plates. mean that CVN would be required at -34C for Zone II (most This is more of a problem in welded members, because weld- of the country). To achieve notch toughness at these lower ing increases the potential for initial flaws and defects that temperatures, the steel would require additional processing, can initiate brittle fracture. Another feature of the specifica- such as normalization. This increased processing would sig- tion requires higher toughness for steels with yield strength nificantly increase the cost of the steel, which is a major part greater than 345 MPa. of total bridge costs. The AASHTO material toughness specifications provide Ultimately, the temperature shift concept was retained, the minimum level of toughness required to sufficiently min- largely out of economic necessity, and the fracture control imize the risk of brittle fracture. The existing AASHTO frac- plan placed a strong emphasis on defect control to prevent ture control plan has generally done a good job of preventing the pop-in-type fracture events. Therefore, the CVN require- fracture failure in bridge structures since design using its pro- ments represented a compromise that allowed the continued visions. A few brittle fractures have still occurred without use of hot-rolled steel rather than normalized steel. noticeable fatigue crack growth in previously designed bridges, however, although the steel met the modern CVN toughness The good fracture-resistance track record of mild steel requirements (i.e., those required by the fracture control plan shows that requiring normalized steel would have been unnec- or general AASHTO toughness requirements). In most such essary. Experience in service and in full-scale experiments has cases, the fracture can be traced back to one or more aspects verified the temperature shift and the adequacy of the notch- of the fracture control plan related to controlling the size of toughness requirements for mild steels. Some premature cracks defects. If all aspects of the fracture control plan are properly and fractures occurred in tied arches fabricated with A514 implemented, the risk of brittle fracture is minimized by the steel. A514 steel is quenched and tempered steel with a mini- current specifications and fracture will be a rare event in ser- mum specified yield stress of 100 ksi. This steel is much dif- vice. However, in the ensuing decades, very few FCBs were

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46 constructed, in part because the fabrication costs were signif- the current AASHTO toughness requirements. The cost pre- icantly greater. mium for higher toughness is lower than it was 30 years ago. High-Performance Steel Consideration of Increasing Toughness Requirements for Bridge Steel The recently developed high-performance steels (HPS) are much tougher but also more expensive than the ordinary Fortunately, today's scrap-based steel typically has improved Grade 250, 345, or 345W steel. The first grade to be fully toughness relative to the steel of the 1970s. This is the result integrated into the AASHTO specifications is the HPS of: (1) a reduction in carbon, (2) a reduction in impurities 485W. Two additional steels are currently being developed, such as phosphorous and sulfur, as well as (3) an increase in an HPS 690 W Cu-Ni and an HPS 345W. Both of these steels alloying elements such as nickel and chromium, which are much tougher than the ordinary Grade 250, 345, and enhance toughness. Nickel and chromium and other beneficial 345W steels. The main reason for developing HPS has been alloys just happen to be in the scrap that is derived primarily to improve weldability, but a large gain in toughness has from automobiles and other sheet metal, which tends to have been a desirable by-product. Development of HPS in the greater alloying than structural steel made from iron ore. United States has been done under a joint research and devel- opment program by FHWA, U.S. Navy, and the American Because steel today has greater toughness than the steels Iron and Steel Institute (A43). of the 1970s is perhaps the best argument for increasing the notch toughness requirements for FCMs. This would allow In contrast to ordinary bridge steel, Grade HPS 485W shows designers to take advantage of the superior toughness char- upper shelf behavior at test temperatures down to approxi- acteristics of HPS (A45). In other words, because the CVN mately -20C. Even at the extremely cold service temperature specifications were essentially just below what was consis- for Zone 3 (LAST = -51C), this steel exhibits toughness in the tently attainable in the 1970s, and because the toughness has upper transition region; without the benefit of the temperature generally improved over time, it is rational to raise the bar shift, the CVN energy still exceeds 150 J. Likewise, high somewhat to ensure that the bridge steel represents the best dynamic and crack arrest toughness can be expected of this practice of today. This could probably be done without forc- steel. Clearly, HPS 485W steel provides a level of toughness ing the mills to use any additional processing and therefore far exceeding the current requirements, which are based on pre- would not significantly affect the cost of steel. Even though the venting brittle fracture. argument can be made that the increased toughness is not absolutely necessary, it is always better to have greater tough- In current research underway at the TurnerFairbank High- ness if there is no cost penalty. way Research Center of FHWA, the fracture toughness of HPS are being determined as a function of temperature, loading rate, If there was no temperature shift, extremely large cracks, and plate thickness (A44,A45). At the same time, the fracture greater than 350 mm, can be tolerated in mild steel. If the notch resistance of fatigue-cracked I-girders fabricated from HPS 485 toughness requirements were increased, the increased tough- is being determined. The test girders are cyclically loaded until ness could also be used to offset the decrease in reliability a crack grows to the desired length; the girder is then cooled to associated with no longer decreasing the stress range for FCMs -34C and subjected to a typical design overload. If the girder in the LRFD Specifications (A2). To make it more palatable, does not fracture, the fatigue crack is grown larger and the same this increase could be combined with some loosening of the overload is reapplied. The HPS 485W steel girder resisted the definition of FCMs that would, for example, allow two-girder full design overload until approximately 50% of the tension bridges that are known to be redundant, as discussed earlier. flange area was lost to fatigue, meaning that the net section of the cracked tension flange had yielded. Research currently underway at FHWA's TurnerFairbank Highway Research Center is studying ways of taking advan- In contrast, the full-scale tests discussed previously (A39, tage of the benefits of higher toughness as in most modern plate A41) on girders made of Grade 345 steel fractured when the net steel, especially HPS. These benefits include: section of the cracked tension flange reached a stress of approx- imately 60% of the yield stress, meaning that the fracture was Elimination of special in-service inspection require- brittle. In contrast, the fracture mode for the Grade HPS 485W ments, such as hands-on or arms-length inspection for indicates a large amount of plastic deformation before failure. fracture-critical structures for HPS. Reduction in the frequency of inspections for the super- The HPS grade steels provide a toughness level that far structure components of HPS bridges. exceeds the minimum requirements cited in Table A2. Elimination of the penalty for structures with low redun- Although HPS 485W costs more than ordinary grade 345W dancy for HPS. steel, the advantages afforded by higher strength more than offset the difference in material costs. This has significantly Ultimately, the greatest benefit might be achieved by pro- altered the economic factors that were considered in setting viding a solid foundation for structural innovation. The cur-

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47 rent U.S. practice of providing a high level of structural redun- Inspection is primarily performed visually. The survey dancy prevents engineers from considering other structural noted that many agencies inspected non-FCBs from the systems that can result in more efficient, lower-cost bridges. ground with binoculars, whereas the FCBs were visually For example, two-girder and tied-arch bridges are rarely built inspected hands on. Nondestructive testing (NDT) is used today in the United States, even though experience has some for FCMs, in particular if there is an indication that is shown that in many situations they are very economical. A clearly not a crack from the visual inspection. NDT methods higher factor of safety against fracture will increase the reli- presently used in service for bridges include but are not lim- ability level of low-redundancy systems, thereby reducing ited to the following: PT, MT, and ultrasonic testing (UT). this barrier to innovation. Radiographic testing and eddy-current testing are usually only used in the shop and therefore are not discussed here. Although higher toughness makes bridges more tolerant to longer cracks, it does not significantly increase fatigue life. Dye-Penetrant Testing Fatigue cracks grow according to a power law; therefore, most of the fatigue life is spent growing the crack while it is PT is used to detect surface discontinuities only. A penetrat- very small. Additional fracture toughness, greater than the ing liquid dye, either visible or fluorescent, is placed on the minimum specified values, will allow the crack to grow surface of the member and will enter any discontinuities. longer before the member fractures. But, at that late stage, the After a period of time, up to 30 min for extremely fine, tight crack is growing so rapidly that relatively few cycles are discontinuities, the excess dye is removed and the area is needed to reach the end of the life. allowed to dry. A developer is then applied, pulling the resid- ual wet dye from the discontinuities as shown in Figure A5. Ultimately, decisions to specify toughness beyond the Penetrant inspection is inexpensive, simple, and easy to minimum required level must be made based on cost. The learn. However, inspectors need to be properly trained in not- current AASHTO steel toughness requirements were devel- ing the difference between real and false indications that oped using the cost factors that existed in the 1970s and con- often occur from a slight weld undercut and other disconti- sidering the state of the art in steel production at that time. nuities that are not significant. Modern steel processing practice has made it more econom- ical to produce high-toughness steels such as A709 HPS 485W. In addition, the Grade 250, 345, and 345W steels pro- Magnetic Particle Testing duced today have typical CVN that far exceed the minimum MT involves the use of magnetic field lines to determine requirements. The current AASHTO specifications for mater- whether surface or near surface cracks exist by the disruption ial toughness may need to be reevaluated to take maximum of the lines. This disruption of lines results from a disconti- advantage of the higher-toughness steels used today. nuity in the member; for example, a crack. The material can either be magnetized through direct magnetization or by plac- INSPECTION AND NONDESTRUCTIVE TESTING ing a magnetic field (indirect magnetization) on the member. Once the field is established, magnetic particles (typically in Periodic in-service inspection provides a final safety net to the form of a powder) are placed on the inspection surface. detect cracks before they grow to a critical size. Because of Discontinuities are exposed when they are trapped in the leak- the repetition of details in a bridge, when one crack is found, age of the magnetic field and the location, shape, and size of it is very likely that similar details may also be cracked. a crack can accurately be determined. Figures A6A8 depict Therefore, the first and most urgent step in planning repairs this process and the required equipment. and retrofits is to thoroughly inspect the bridge for other cracks, usually visually but often followed up with nonde- MT can be conducted very quickly and, compared with structive evaluation such as magnetic-particle testing (MT) other NDT methods, it is relatively cost-effective in terms of or dye-penetrant testing (PT), especially for FCMs. The focus of inspection for fatigue cracks should be on details similar to the one that cracked on elements of the bridge in tension, with a priority for elements with high live-load stress ranges and FCMs (A46A48). Elements for which the applied stress remains in compres- sion need not be inspected closely for fatigue cracks, because complete fracture is not possible. Owing to welding residual stresses, a crack can still occur in a structural element that undergoes cyclic loading even if the applied stress remains in compression. However, these cracks will usually arrest as they grow away from the welds as the tensile residual stress field either decreases or is relieved by the cracking (A4). FIGURE A5 Example of indication from dye penetrant inspection.

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48 equipment and procedures. In contrast to PT, MT can reveal shallow cracks below the surface, is very accurate, requires less time, and may be more economical after the equipment is obtained. This procedure is favored among many inspectors. Ultrasonic Testing UT inspection is another commonly used NDT method in practice. By using high-frequency sound waves, surface and subsurface discontinuities can be detected. As the sound waves travel through the material and reflect back, the presence and location of any discontinuities, which also cause reflections, in the member can be detected. This information is then dis- played on a cathode ray tube screen for interpretation. Advantages in using this type of test are the ability to FIGURE A6 Magnetic particle testing--Placement of magnetic detect small internal discontinuities, accuracy, and nearly field. instantaneous test results. The primary disadvantage to this test is that highly trained and experienced technicians are needed to operate and accurately interpret this type of test results. The international fabrication scanning tour noted that automated UT, providing a permanent record, is often used outside the United States. Because small, possibly innocuous, discontinuities can be detected with UT, acceptance criteria are presented in AWS D1.5. These acceptance criteria are workmanship standards; that is, they represent the typical quality level easily achiev- able by good welders. The AWS D1.5 UT acceptance crite- ria are not based on the effect that the rejectable discontinu- ities might have on the resistance to fatigue and fracture; they are typically more strict than necessary. The AWS standards are for new fabrication and were not intended for existing structures. It is a misapplication to FIGURE A7 Magnetic particle testing--Application of magnetic apply these workmanship standards to an evaluation of exist- particles. ing bridges. If there is no impact on fatigue and fracture, an owner will be far more reluctant to take out of service or repair (at owner expense) a structure that is found to have poor workmanship than the owner would be when the com- ponent is still in the fabrication shop and the repairs would be at the fabricator's expense. Notwithstanding that it is a misapplication, the AWS D1.5 criteria have frequently been used to assess the UT of butt welds in service. On many occasions, an indication with a rejectable decibel (dB) rating will be cored out of a large groove weld for destructive examination and characterization of the actual flaw. Figure A9 shows the results of many of these investigations, as conducted by Dr. Eric Kaufmann at Lehigh University's Advanced Technology for Large Struc- tural Systems (ATLSS) Center, plotted in terms of the dB rat- ing versus the actual flaw size. Also shown are the D1.5 rejec- tion limits for various thickness butt welds. It can be seen that the AWS criteria are very conservative. For typical plating thickness, the AWS criteria will reliably screen out defects of FIGURE A8 Magnetic particle testing--Indication. only a few millimeters in width.

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49 Surface Treatments 10 Weld toe surface treatments include grinding, gas tungsten 5 Reject for t < 18 mm (70o class A) arc or plasma remelting of the weld toe, and impact treat- 0 Defect ments. These techniques can be used as "weld improvement" Rating Reject for t < 64 mm -5 retrofit methods; that is, for increasing the fatigue strength of (db) -10 Reject for t < 203 mm uncracked welds. With any of these treatments, the improve- ment in fatigue strength can be attributed to one or a combi- -15 nation of the following: 0 2 4 6 8 10 12 Improvements in the weld geometry and corresponding Measured Defect Width (2a) (mm) reduction in the stress concentration, Elimination of some of the more severe discontinuities FIGURE A9 Actual defect size from destructive from which the fatigue cracks propagate, or examinations of cores from welds with UT indications of various dB rating. (Data from Lehigh University.) Reduction of tensile residual stress or the introduction of compressive residual stress (A49,A50). The easiest and lowest cost of these treatments is hammer Established fracture mechanics principles can be used to peening, which is very effective and commonly used. Some define acceptable initial crack sizes that will not propagate to of the methods, including hammer peening, can also be used the critical size in the lifetime of a structure (A15,A32). These for the repair of shallow surface cracks up to 3 mm deep. types of calculations are referred to as "fitness-for-purpose" calculations, indicating that although a component may have A relatively new process, known as Ultrasonic Impact rejectable discontinuity, it can be proven that the component Treatment, has been the subject of several recent studies is fit for its defined purpose (lifetime and anticipated loading). (A51,A52). The process, developed in Russia, is similar to air hammer peening, but applies the treatment at a very high fre- RETROFIT METHODS quency, up to 35 kHz. This technique uses sound waves to excite a peening device that introduces compression into the This section presents commonly used repair and retrofit tech- steel at the toe of fillet welds. Fatigue testing done on treated niques for fatigue-critical details, as well as retrofit tech- and untreated specimens cut from the plate girder concluded niques to improve the redundancy of fracture-critical bridges that, for the conditions tested, Ultrasonic Impact Treatment (FCBs). A distinction is made between a repair and a retro- altered the performance of a Category C detail by imparting fit: a repair is intended to arrest the propagation of a fatigue to it the fatigue strength of a Category B detail. Further crack, whereas a retrofit is intended to either (1) upgrade the research is being conducted to address the effect Ultrasonic fatigue resistance and prevent the occurrence of fatigue Impact Treatment will have, if any, on the fatigue threshold cracking, or (2) create an alternative load path in the event of of details and how different fatigue stress ranges, welding fracture to make an FCM into a non-FCM. processes, and quality control affect the results. Issues related to repair and retrofit of fatigue cracks will Once either treatment is applied to the welds that have be discussed first followed by some discussion related to already been in service, the remaining fatigue life is at least retrofit of FCBs. One of the best sources for information as good as the life of the original detail when it was new. In related to retrofit strategies is Fatigue and Fracture in Steel other words, there is no remaining effect of prior fatigue load- Bridges by John W. Fisher (A13). Although out of print, it ing cycles. In most cases, these treatments result in fatigue remains an excellent source for material on retrofitting fatigue- strength of the treated detail that is at least one fatigue "cat- or fracture-damaged bridges. egory" greater than the original detail; that is, the next great- est SN curve in the AASHTO set of SN curves can be used Repair and Retrofit of Fatigue Cracks to predict the residual life of the repaired detail. Surface treat- ments only affect the weld toes; therefore, fatigue cracks may Many different methods are used for the repair of fatigue still develop from the weld roots. cracks and retrofit of fatigue-prone details. The choice of method depends on the circumstances of the fatigue cracking Hole Drilling and may also depend on the availability of certain skills and tools from local contractors who would perform the repairs. Hole drilling is perhaps the most widely used repair method Repair and retrofit techniques can be placed in three major for fatigue cracks or retrofit method for fatigue-critical details. categories: (1) surface treatments, (2) repair of through- It is often used as a temporary measure to arrest a propagating thickness cracks, and (3) modification of the connection or crack, followed eventually by more extensive repairs. It is rare the global structure to reduce the cause of cracking. that any repair scheme such as repair welding or modifying

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50 the detail does not begin with drilling the crack tips. For retrofit, hole drilling is often used to isolate a detail or to intercept potential cracks before they can propagate far into main elements. By properly tensioning a high-strength bolt in the hole it can be considered a Category B detail. A hole by itself is basically a Category D detail. For repairs, the hole drilling method requires placing a hole at the tip of the crack, essentially blunting the tip of the crack, thus removing the high-stress concentration associ- ated with the sharp tip. However, the hole needs to be a spe- cific diameter to be successful in arresting the crack. Typi- cally, a hole diameter of 4 in. (200 mm) is recommended because this has proven to be effective. When a more refined estimate of the required hole size is necessary, relationships have been developed to define the size of the hole needed to arrest the crack (A53,A54). Appropriate checks on the net section capacity of the member should be made. The fatigue resistance of a hole can also be increased by the cold expansion of the hole, which upsets the material around the perimeter of the hole and introduces beneficial compressive residual stress around the hole. This technique is widely used in aluminum airframes. Once the hole is FIGURE A10 Bolted doubler plate repair. Dotted line represents drilled, a tapered mandrel (also referred to as a drift pin) crack line beneath doubler plate and circle is the hole drilled at slightly larger than the hole can be forced through the hole crack tip to intercept further growth. by hitting the pin with a hammer. As the pin passes through the hole, the hole plastically deforms creating the compres- sive field around the hole. This method will not work when will have adequate fatigue resistance, doubler plates can be the hole used contains a crack. If the hole has a crack enter- added after the weld repair to decrease the stress range. ing it, the hole is forced open by the pin as the pin is driven into the hole because the cracked edge is compliant (flexible) The one problem with this repair is the alignment of the two and the hole does not provide sufficient constraint to induce sides of the crack before the weld repair. As can be seen in Fig- the compressive stresses at the edge. However, if the hole is drilled ahead of the crack tip, the hole may be cold expanded. Adding Doubler and Splice Plates Another technique that can be used to repair through-thickness cracks is by the addition of doubler plates, or doublers. Dou- bler plates add material to the cross section to either increase or make up for the cracked cross-sectional area. Doubler plates may be bolted (Figure A10) or welded (Figure A11). From a fatigue-resistance standpoint, bolted doublers are always bet- ter than welded ones, because a high-strength bolted connec- tion can be considered an AASHTO Category B detail, whereas a welded connection will be Category E or worse. It is therefore usually recommended that only bolted doubler plates be used for permanent repair or retrofit on bridges. The philosophy of doublers for fatigue crack repair is to add cross-sectional area, which in turn reduces stress ranges. For instance, if a fatigue crack grows across the full depth of the bridge girder, there are two ways that it can be repaired. First, a vee and weld repair can be specified, but the base metal that is weld repaired will have at best a Category D FIGURE A11 Welded doubler plate detail. (Note that corners fatigue resistance (A54A56). To ensure that the weld repair should be rounded.)

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51 ure A12, the cracked surface usually develops buckles, making tilevered truss that has a very long suspended center span. their alignment difficult. In this case, the second option would Stainless steel rods were added to supplement the primary be to use thicker doublers and bolt them to the girder. The hanger members in the event that the existing hanger truss thicker plates add enough cross section assuming the crack will member were to fracture. The members are preloaded so that not be weld repaired, and bolting them together then ensures the additional rods carry a portion of the dead load and so that that any buckles can be straightened out. This technique is par- if a hanger were to fracture, there would be minimal "snap" ticularly useful when a full-depth crack forms in a bridge girder. as the full load was dynamically transferred to the rods. Sim- The doubler plates are then meant to make up for the lost cross ilar systems have been incorporated to other tension members section from the crack. Doublers are also typically used to in other trusses such as diagonals and chords. restore a section that has been heavily damaged by corrosion. Another related retrofit technique is to string high- Retrofit of Fracture-Critical Members to Make strength strands along the bottom face of the bottom flange Them non-Fracture-Critical Members of a fracture-critical plate girder. The strands are then pre- loaded and a portion of the dead load transferred to the newly There are a few methods that have been developed that were added strands. Thus, in the event the existing tension flange identified in the literature to improve the redundancy of fractures, there are additional tension members (strands) with FCBs. Some of these were illustrated in the main body of the sufficient capacity available to carry both dead and live report. Others will be summarized briefly here. loads. In addition, depending on the level of post-tensioning applied, the dead-load stresses (as well as live-load stresses) can be reduced in the existing bottom flange. Prestressing Strands On some bridges, the addition of prestressing strands or rods Another approach that used post-tensioning was on the has been used to supplement tension members. An example Hazard, Woodhead, Dunlavy, and Mandell Street tied arch is the Girard Point Bridge that carries I-95 over the Schuylkill bridges in Houston, Texas. In these bridges, the 2 ft 2 ft tie River near Philadelphia. The bridge is a double-deck can- girder was internally post-tensioned and encased in concrete. (As part of the strategy to improve redundancy the tie was not encased in concrete.) Four post-tensioning strands are in each tie. However, because the tie is encased in concrete, it is inac- cessible for future inspection. These recently built steel tied arch bridges span 224 ft (68 m) over the freeway and carry two lanes of traffic, two bicycle lanes, a utility parapet in each direction, and sidewalks outside of each arch for a total width of 60 ft (18 m). One of these spans is shown in Figure A13. Note the shallow tie girder. Post-tensioning of the arch tie provided redundancy and virtually eliminated tension in the tie, which allayed concerns about the history of problems with tie beams on other tied arch bridges, but necessitated passing the arch tie flanges through the junction with the arch rib. FIGURE A12 Full-girder-depth fatigue crack of Lafayette Street Bridge in St. Paul, Minnesota. FIGURE A13 Post-tension tied arch bridge in Houston, Texas.

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52 Bolted Redundancy Plates Stringer note Another technique that has been used on some bridges is shear studs the addition of bolted redundancy plates. This retrofit con- sists of bolting plates or angles to existing tension mem- bers. The primary function is to provide an additional com- ponent(s) so that if the tension flange of an existing member were to fail at some location along the length, the added component would assume the full dead and live loads of the tension flange. For example, the two-girder approach spans on the Poplar Street Bridge in East St. Louis, Shear studs Missouri, have been retrofit by bolting thick HPS plates on top chord that would take the place of the tension flange along the web just above the tension flange, as shown in Figure 15 in chapter two of this report (page 17). Because the added components are bolted to the existing FIGURE A14 Shear studs being installed on top chords, member, there is no direct path for the fracture to travel into floorbeams, and stringers on the SR-33 bridge near Easton, the added component. Hence, the member becomes inter- Pennsylvania. nally redundant as with a riveted built-up member. For exam- ple, in a riveted tension flange comprised of several plates it has been observed that the cover plate can fully fracture, Controlled load testing and long-term monitoring of the although the other elements of the member remain and the bridge confirms that there is excellent load distribution member continues to take the entire load. When a retrofit between trusses and that stresses in the upper chord in the plate is added to a bridge with no live load, it also reduces the negative moment regions are very small owing to the com- live-load stress range in the existing member. posite action (A56). Hence, it is believed that the choice to make the deck composite with the upper chord added sub- The technique has also been used in new construction on stantial redundancy. a large truss bridge that carries SR-33 over the Lehigh River near Easton, Pennsylvania. On this new bridge, redundancy Installation of Additional Girders plates were bolted alongside of selected tension chords that were identified to be critical, as shown in Figure 13 in chap- This retrofit technique is only applied when a structure is to ter two of the report (page 16). The plates were fully spliced be widened. In cases where the existing structure is function- at the panel points and nominally connected to the member ally obsolete and additional or wider lanes are required, an along the length. An advantage of redundancy plates used in additional exterior girder is sometimes added. If the new the design is that the components share both dead and live girder is adequately attached to the existing girders, full load loads throughout the life of the bridge. sharing can be realized. Thus, a two-girder bridge can become a four-girder bridge and be removed from the list of FCBs in Although these techniques add internal member redun- an owner's inventory. It is emphasized that the new girder dancy, they do not add overall structural load path redun- must be sufficiently attached to the existing girder and deck. dancy. In other words, in the unlikely event that the entire lower chord failed only one lower chord remains. This technique was employed on the Pennsylvania Turn- pike near the Valley Forge Interchange northwest of Philadel- Use of Composite Construction phia. In this example, a two-girder bridge was widened by adding parallel girders adjacent to the existing riveted girders. The SR-33 Bridge near Easton, Pennsylvania, incorporated In effect, a two-girder cross section was converted into a four- an additional measure to increase redundancy. The top chord girder cross section. Although the girder spacings are unequal is fully composite with the concrete deck (see Figure A14). and the distance between the two original girders may still be Traditionally, in truss bridges, only the floorbeams and considerable, the structure as a system should not be consid- stringers are made composite with the deck. The deck is cast- ered fracture critical as long as the connections between the in-place reinforced concrete supported by steel stringers and girders and to the concrete deck are adequate. transverse floorbeams. It is the only composite truss in the state of Pennsylvania and possibly the United States. The The additional girders can also be used to retrofit cross structure is a four-span continuous haunched steel deck truss girders. For example, two girders were used to provide that is composite with the reinforced concrete deck. The redundancy to existing steel cross girders carrying multi- main river span is 181 m and the depth of the trusses varies girder composite spans, as shown in Figure A15. The bridge from 11 m to 22 m. carries I-95 just north of Philadelphia. The retrofit required

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53 Existing cross girder New steel support columns (typ.) Newly added cross girders on each side of cross girder FIGURE A15 Additional support girders are used to provide redundancy to steel cross girders on I-95 bridge north of Philadelphia, Pennsylvania. the addition of new steel support columns to carry the loads The ability of the structure to behave as a continuous mul- from the new cross girders to the footings. The retrofit was tispan bridge, primarily in the negative moment regions, installed as a preemptive strategy, because no problems with must be adequately checked. During construction, either the cross girders have been observed. false work or strong backs are required to ensure that the bridge is stable. The process can be completed with a live load on the bridge. Figure A17 illustrates a two-girder bridge Pin and Hanger Retrofits where the pin and hanger were removed and replaced. There are two common strategies used to retrofit pin and hanger bridges. Both will be discussed here. It is important DEVELOPMENT OF FRACTURE CONTROL PLAN to note that although this system improves the redundancy of the given pin and hanger assembly, it does not necessarily This section reviews the history of the development of the add load path redundancy to the entire structure. However, fracture toughness requirements for steel and weld filler for the approaches that actually remove the pin and hanger metal, which are a central part of the bridge fracture control and replace the detail with a full moment connection, the case plan contained in D1.5. The requirements appear to have could be made that, if sufficient diaphragms are present, served their purpose; that is, there have been no catastrophic there may be sufficient alternate load paths. Addition of Supplemental "Catcher" Systems This form of retrofit is typically used on pin and hanger sys- tems. In the typical application, an additional group of com- ponents are added to "catch" the suspended girder should the existing pin and hanger system fail. A typical installation of this system is shown in Figure A16. Removal of Pin and Hanger Assembly In this approach, the entire pin and hanger assembly is "Catcher" beam removed and replaced with a new short section of girder that is attached to existing portions of the girders with full moment splices. The girders are then made continuous for live load and even some proportion of dead load. Field instru- mentation conducted on the bridge in Figure A17 confirmed that after the retrofit, the bridge behaved as a typical contin- FIGURE A16 Catcher system as used on a typical pin and uous multispan bridge (A57). hanger bridge. (Courtesy: Modjeski and Masters, Inc.)

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54 Factor Design, 3rd ed., American Institute of Steel Construction, Chicago, Ill., 1999. A5. "Steel Structures," AREMA Manual for Railway Engi- neering, American Railway Engineering and Mainte- nance of Way Association, Landover, Md., 2002. A6. Structural Welding Code--Steel, ANSI/AWS D1.1- 02, American Welding Society, Miami, Fla., 2002. A7. Limit States Design of Steel Structures, CSA S16- 2001, Canadian Standards Association, Toronto, ON, Canada, 2001. A8. Fisher, J.W., K.H. Frank, M.A. Hirt, and B.M. McNamee, NCHRP Report 102: Effect of Weldments Location of former on the Fatigue Strength of Steel Beams, Highway pin and hanger Research Board, National Research Council, Wash- ington, D.C., 1970, 114 pp. A9. Fisher, J.W., P.A. Albrecht, B.T. Yen, D.J. Klinger- FIGURE A17 Typical complete girder splice installed to replace man, and B.M. McNamee, NCHRP Report 147: a pin and hanger connection in a two-girder bridge. Fatigue Strength of Steel Beams with Welded Stiffen- ers and Attachments, Transportation Research Board, National Research Council, Washington, D.C., 1974, fracture problems with bridges since their implementation. 85 pp. However, many things have changed since the development A10. Keating, P.B. and J.W. Fisher, NCHRP Report 286: of these specifications. There are fundamental differences in Evaluation of Fatigue Tests and Design Criteria the steel and the way it is produced. Some of these differ- on Welded Details, Transportation Research Board, ences may have some impact on the temperature shift and National Research Council, Washington, D.C., Sep. other assumptions in the original development of the tough- 1986, 66 pp. ness requirements. A11. Petershagen, H. and W. Zwick, Fatigue Strength of Butt Welds Made by Different Welding Processes, The difference between the fracture control plan provisions IIW-Document XIII-1048-82, International Institute and the provisions for non-FCMs elsewhere in AASHTO/ of Welding, West Germany, 1982. AWS D1.5 is primarily that there are more strict fabrication A12. Petershagen, H., "The Influence of Undercut on and shop-inspection requirements to control weld flaws and the Fatigue Strength of Welds--A Literature other crack-like defects. For example, both radiographic test- Survey," Welding in the World, Vol. 28, No. 7/8, 1990, ing and ultrasonic testing are required on all groove welds for pp. 2936. fracture control elements. In addition, CVN requirements for A13. Fisher, J.W., Fatigue and Fracture in Steel Bridges, welds and base metal are increased for fracture control ele- John Wiley and Sons, New York, N.Y., 1984, 336 pp. ments. The provisions result in an even lower probability of A14. Anderson, T. L., Fracture Mechanics--Fundamentals brittle fracture in new FCMs than for typical non-FCMs. and Applications, 2nd ed., CRC Press, Boca Raton, Fla., 1995, 88 pp. REFERENCES A15. Barsom, J.M. and S.T. Rolfe, Fracture and Fatigue Control in Structures: Applications of Fracture A1. Task Committee on Redundancy of Flexural Systems Mechanics, 3rd ed., American Society for Testing and of the ASCEAASHTO Committee on Flexural Mem- Materials, West Conshohocken, Pa., 1999. bers of the Committee on Metals of the Structural A16. Broek, D., Elementary Fracture Mechanics, 4th ed., Division, "State-of-the Art Report on Redundant Martinis Nijhoff Publishers, Dordrecht, the Nether- Bridge Systems," Journal of Structural Engineering, lands, 1987. Vol. 111, No. 12, Dec. 1985. A17. Kober, W., E.J. Dexter, E.J. Kaufmann, B.T. Yen, and A2. AASHTO LRFD Bridge Design Specifications, 3rd ed., J.W. Fisher, "The Effect of Welding Discontinuities American Association of State Highway and Trans- on the Variability of Fatigue Life," Fracture Mechan- portation Officials, Washington, D.C., 2004. ics, Vol. 25, ASTM STP 1220, F. Erdogan and R.J. A3. Fisher, J.W., "The Evolution of Fatigue Resistant Steel Hartranft, Eds., American Society for Testing and Bridges," 1997 Distinguished Lectureship, Paper No. Materials, Philadelphia, Pa., 1994. 971520, 76th Annual Meeting of the Transportation A18. Guide for Fatigue Strength Assessment of Tankers, Research Board, Washington, D.C., Jan. 1216, 1997, American Bureau of Shipping, New York, N.Y., June pp. 122. 1992. A4. LRFD Specification for Structural Steel Buildings, A19. Fatigue Design Guidance for Steel Welded Joints in Manual of Steel Construction: Load and Resistance Offshore Structures, UK Health & Safety Executive (formerly the UK Department of Energy), Her

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55 Majesty's Stationery Office, London, United King- A34. Dexter, R.J., W.J. Wright, and J.W. Fisher, "Fatigue dom, 1984. and Fracture of Steel Girders," Journal of Bridge A20. Roberts, R., et al., Corrosion Fatigue of Bridge Steels, Engineering, Vol. 9, No. 3, May/June 2004, pp. Vols. 13, Reports FHWA/RD-86/165, 166, and 167, 278286. Federal Highway Administration, Washington, D.C., A35. Dexter, R.J., "Fatigue and Fracture," The Structural May 1986, 44, 168. 564 pp. Engineering Handbook, 2nd ed., E.M. Lui, Ed., CRC A21. Outt, J.M.M., J.W. Fisher, and B.T. Yen, Fatigue Press, Boca Raton, Fla., 2004. Strength of Weathered and Deteriorated Riveted A36. "State Cites Defective Steel in Bryte Bend Failure," Members, Report DOT/OST/P-34/85/016, Federal Engineering News Record, Vol. 185, No. 8, Aug. 20, Highway Administration, Department of Transporta- 1970. tion, Washington, D.C., Oct. 1984, 138 pp. A37. Vehovar, L., "Hydrogen-Assisted Stress-Corrosion of A22. Albrecht, P. and C. Shabshab, "Fatigue Strength of Prestressing Wires in a Motorway Viaduct," Engineer- Weathered Rolled Beam Made of A588 Steel," Jour- ing Failure Analysis, Vol. 5, No. 1, 1998, pp. 2127. nal of Materials in Civil Engineering, Vol. 6, No. 3, A38. Kulak, G.L., J.W. Fisher, and J.H.A. Struick, Guide to 1994, pp. 407428. Design Criteria for Bolted and Riveted Joints, 2nd ed., A23. Fisher, J.W., A. Nussbaumer, P.B. Keating, and B.T. Prentice Hall, Englewood Cliffs, N.J., 1987, 352 pp. Yen, NCHRP Report 354: Resistance of Welded A39. Roberts, R. and G.V. Krishna, "Fracture Behavior of Details Under Variable Amplitude Long-Life Fatigue A36 Bridge Steels," Report FHWA-RD-77-156, Fed- Loading, Transportation Research Board, National eral Highway Administration, Washington, D.C., Research Council, Washington, D.C., 1993, 38 pp. 1977, 59 pp. A24. Miner, M.A., "Cumulative Damage in Fatigue," Jour- A40. Roberts, R., G.R. Irwin, G.V. Krishna, and B.T. Yen, nal of Applied Mechanics, Vol. 12, 1945, p. A-159. "Fracture Toughness of Bridge Steels--Phase II," A25. Guide Specifications for the Fatigue Evaluation of Report FHWA-RD-74-59, Federal Highway Adminis- Existing Steel Bridges, American Association of State tration, Washington, D.C., 1974, 418 pp. Highway and Transportation Officials, Washington, A41. Schilling, C.G., K.H. Klippstein, J.M. Barsom, S.R. D.C., 1990. Novak, and G.T. Blake, Low Temperature Tests of A26. Moses, F., C.G. Schilling, and K.S. Raju, NCHRP Simulated Bridge Members, Report No. 97.021- Report 299: Fatigue Evaluation Procedures for Steel 001(3), American Iron and Steel Institute, Washing- Bridges, Transportation Research Board, National ton, D.C., 1972. Research Council, Washington, D.C., 1987, 100 pp. A42. Hartbower, C.E., "Reliability of the AASHTO Tem- A27. Dexter, R.J. and J.W. Fisher, "Fatigue and Fracture," perature Shift in Material Toughness Testing," Struc- In Handbook of Bridge Engineering, W.F. Chen, Ed., tural Engineering Series No. 7, Federal Highway CRC Press, Boca Raton, Fla., 1999. Administration, Washington, D.C., 1979. A28. Youngberg, C.J., R.J. Dexter, and P.M. Bergson, A43. Wright, W.J., "High Performance Steel--Research to Fatigue Evaluation of Steel Box-Girder Pier Caps: Practice," Public Roads, Vol. 60, No. 4, 1997, pp. 3438. Bridge 69832, Report MN/RC 2003-18, Minnesota A44. Wright, W.J., "Fatigue Strength and Fracture Resis- Department of Transportation, St. Paul, Mar. 2004, tance of HPS-485W High Performance Steel--Work 67 pp. in Progress," Paper No. T103-6, Proceedings, Struc- A29. Wright, W.J., J.W. Fisher, and B. Sivakumar, "Hoan tural Engineers World Congress, Elsevier, 1998. Bridge Failure Investigation," Federal Highway A45. Wright, W.J., H. Tjiang, J. Hartman, and P. Albrecht, Administration, Washington, D.C., 2001. "Fracture Resistance of Modern Bridge Steels," Pro- A30. Kaufman, E.J., R.J. Connor, and J.W. Fisher, "Failure ceedings, ASCE Structures Congress, Philadelphia, Analysis of the US 422 Girder Fracture--Final Pa., 2000. Report," ATLSS Report No. 04-21, Center for A46. Yen, B.T., T. Huang, L.-Y. Lai, and J.W. Fisher, Advanced Technology for Large Structural Systems, Manual for Inspecting Bridges for Fatigue Damage Lehigh University, Bethlehem, Pa., Oct. 2004. Conditions: Final Report, Report FHWA-PA-89-022 A31. Connor, R.J., E.J. Kaufmann, J. Jin, and C.H. Davies, + 85-02, Fritz Engineering Laboratory, Bethlehem, "Forensic Investigation of the SR422 Over the Pa., 1990, 174 pp. Schuylkill River Girder Fracture," Proceedings of the A47. Demers, C. and J.W. Fisher, Fatigue Cracking of Steel Twenty-First International Bridge Conference, Pitts- Bridge Structures, Vol. I: A Survey of Localized Crack- burgh, Pa., June 1416, 2004. ing in Steel Bridges--1981 to 1988, Report FHWA- A32. "Guide on Methods for Assessing the Acceptability of RD-89-166, also, Vol. II: A Commentary and Guide for Flaws in Metallic Structures," BS 7910, British Stan- Design, Evaluation, and Investigating Cracking, Report dards Institute, London, 1999. FHWA-RD-89-167, Federal Highway Administration, A33. Dexter, R.J. and J.W. Fisher, "Fatigue and Fracture," McLean, Va., Mar. 1990. In Steel Design Handbook, LRFD Method, A.R. Tam- A48. Harland, J.W., R.L. Purvis, D.R. Graber, P. Albrecht, boli, Ed., McGrawHill, New York, N.Y., 1997. and T.S. Flournoy, Inspection of Fracture Critical

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56 Bridge Members, supplement to the Bridge Inspec- A53. Fisher, J., B.M. Barthelemy, D.R. Mertz, and J.A. tor's Training Manual, final report, Report FHWA-IP- Edinger, NCHRP Report 227: Fatigue Behavior of 86-26 (PB87-163440), Federal Highway Administra- Full-Scale Welded Bridge Attachments, Transporta- tion, McLean, Va., Sept. 1986, 232 pp. tion Research Board, National Research Council, A49. Braid, J., R. Bell, and D. Militaru, "Fatigue Life of as- Washington, D.C., 1980, 47 pp. Welded, Repaired, and Hammer-Peened Joints in A54. Dexter, R., R. Fitzpatrick, and D. St. Peter, "Fatigue High-Strength Structural Steel," Welding in the Strength and Adequacy of Fatigue Crack Repairs," World/Le Soudage le Monde, Vol. 39, No. 5, 1998, Report SSC-425, Ship Structure Committee, Wash- pp. 248261. ington, D.C., May 2003. A50. Harrison, J.D., "Further Techniques for Improving the A55. Gregory, E.N., G. Slater, and C.C. Woodley, NCHRP Fatigue Strength of Welded Joints," British Welding Report 321: Welded Repair of Cracks in Steel Bridge Journal, Vol. 13, No. 11, 1966. Members, Transportation Research Board, National A51. Tryfyakov, V., P. Mikheev, Y. Kudryavtsev, and D. Research Council, Washington, D.C., 1989, 52 pp. Reznik, "Ultrasonic Impact Peening Treatments of A56. Dexter, R.J. and B.A. Kelly, "Research on Repair and Welds and Its Effect on Fatigue Resistance in Air and Improvement Methods," International Conference on Seawater," 25th Annual Offshore Technology Confer- Performance of Dynamically Loaded Welded Struc- ence, Houston, Tex., May 36, 1993. tures, Proceedings of the IIW 50th Annual Assembly A52. Roy, S., J. Fisher, and B. Yen, "Fatigue Resistance Conference, San Francisco, Calif., July 1319, 1997, of Welded Details Enhanced by Ultrasonic Impact pp. 273285. Treatment (UIT)," Proceedings of the Eleventh A57. Frank, K.H. and C.F. Galambos, "Application of International Offshore and Polar Engineering Con- Fracture Mechanics to Analysis of Bridge Failure," ference, Stavenger, Norway, June 1722, 2001, Proceedings, Specialty Conference on Safety and Re- pp. 309313. liability of Metal Structures, 1972.