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Rock-Socketed Shafts for Highway Structure Foundations (2006)

Chapter: Chapter Two - Site and Geomaterial Characterization

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Suggested Citation:"Chapter Two - Site and Geomaterial Characterization." National Academies of Sciences, Engineering, and Medicine. 2006. Rock-Socketed Shafts for Highway Structure Foundations. Washington, DC: The National Academies Press. doi: 10.17226/13975.
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Suggested Citation:"Chapter Two - Site and Geomaterial Characterization." National Academies of Sciences, Engineering, and Medicine. 2006. Rock-Socketed Shafts for Highway Structure Foundations. Washington, DC: The National Academies Press. doi: 10.17226/13975.
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Suggested Citation:"Chapter Two - Site and Geomaterial Characterization." National Academies of Sciences, Engineering, and Medicine. 2006. Rock-Socketed Shafts for Highway Structure Foundations. Washington, DC: The National Academies Press. doi: 10.17226/13975.
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Suggested Citation:"Chapter Two - Site and Geomaterial Characterization." National Academies of Sciences, Engineering, and Medicine. 2006. Rock-Socketed Shafts for Highway Structure Foundations. Washington, DC: The National Academies Press. doi: 10.17226/13975.
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Suggested Citation:"Chapter Two - Site and Geomaterial Characterization." National Academies of Sciences, Engineering, and Medicine. 2006. Rock-Socketed Shafts for Highway Structure Foundations. Washington, DC: The National Academies Press. doi: 10.17226/13975.
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Suggested Citation:"Chapter Two - Site and Geomaterial Characterization." National Academies of Sciences, Engineering, and Medicine. 2006. Rock-Socketed Shafts for Highway Structure Foundations. Washington, DC: The National Academies Press. doi: 10.17226/13975.
×
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Suggested Citation:"Chapter Two - Site and Geomaterial Characterization." National Academies of Sciences, Engineering, and Medicine. 2006. Rock-Socketed Shafts for Highway Structure Foundations. Washington, DC: The National Academies Press. doi: 10.17226/13975.
×
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Suggested Citation:"Chapter Two - Site and Geomaterial Characterization." National Academies of Sciences, Engineering, and Medicine. 2006. Rock-Socketed Shafts for Highway Structure Foundations. Washington, DC: The National Academies Press. doi: 10.17226/13975.
×
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Suggested Citation:"Chapter Two - Site and Geomaterial Characterization." National Academies of Sciences, Engineering, and Medicine. 2006. Rock-Socketed Shafts for Highway Structure Foundations. Washington, DC: The National Academies Press. doi: 10.17226/13975.
×
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Suggested Citation:"Chapter Two - Site and Geomaterial Characterization." National Academies of Sciences, Engineering, and Medicine. 2006. Rock-Socketed Shafts for Highway Structure Foundations. Washington, DC: The National Academies Press. doi: 10.17226/13975.
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Suggested Citation:"Chapter Two - Site and Geomaterial Characterization." National Academies of Sciences, Engineering, and Medicine. 2006. Rock-Socketed Shafts for Highway Structure Foundations. Washington, DC: The National Academies Press. doi: 10.17226/13975.
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Suggested Citation:"Chapter Two - Site and Geomaterial Characterization." National Academies of Sciences, Engineering, and Medicine. 2006. Rock-Socketed Shafts for Highway Structure Foundations. Washington, DC: The National Academies Press. doi: 10.17226/13975.
×
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Suggested Citation:"Chapter Two - Site and Geomaterial Characterization." National Academies of Sciences, Engineering, and Medicine. 2006. Rock-Socketed Shafts for Highway Structure Foundations. Washington, DC: The National Academies Press. doi: 10.17226/13975.
×
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Suggested Citation:"Chapter Two - Site and Geomaterial Characterization." National Academies of Sciences, Engineering, and Medicine. 2006. Rock-Socketed Shafts for Highway Structure Foundations. Washington, DC: The National Academies Press. doi: 10.17226/13975.
×
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Suggested Citation:"Chapter Two - Site and Geomaterial Characterization." National Academies of Sciences, Engineering, and Medicine. 2006. Rock-Socketed Shafts for Highway Structure Foundations. Washington, DC: The National Academies Press. doi: 10.17226/13975.
×
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Suggested Citation:"Chapter Two - Site and Geomaterial Characterization." National Academies of Sciences, Engineering, and Medicine. 2006. Rock-Socketed Shafts for Highway Structure Foundations. Washington, DC: The National Academies Press. doi: 10.17226/13975.
×
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Suggested Citation:"Chapter Two - Site and Geomaterial Characterization." National Academies of Sciences, Engineering, and Medicine. 2006. Rock-Socketed Shafts for Highway Structure Foundations. Washington, DC: The National Academies Press. doi: 10.17226/13975.
×
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Suggested Citation:"Chapter Two - Site and Geomaterial Characterization." National Academies of Sciences, Engineering, and Medicine. 2006. Rock-Socketed Shafts for Highway Structure Foundations. Washington, DC: The National Academies Press. doi: 10.17226/13975.
×
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Suggested Citation:"Chapter Two - Site and Geomaterial Characterization." National Academies of Sciences, Engineering, and Medicine. 2006. Rock-Socketed Shafts for Highway Structure Foundations. Washington, DC: The National Academies Press. doi: 10.17226/13975.
×
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Suggested Citation:"Chapter Two - Site and Geomaterial Characterization." National Academies of Sciences, Engineering, and Medicine. 2006. Rock-Socketed Shafts for Highway Structure Foundations. Washington, DC: The National Academies Press. doi: 10.17226/13975.
×
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Suggested Citation:"Chapter Two - Site and Geomaterial Characterization." National Academies of Sciences, Engineering, and Medicine. 2006. Rock-Socketed Shafts for Highway Structure Foundations. Washington, DC: The National Academies Press. doi: 10.17226/13975.
×
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Suggested Citation:"Chapter Two - Site and Geomaterial Characterization." National Academies of Sciences, Engineering, and Medicine. 2006. Rock-Socketed Shafts for Highway Structure Foundations. Washington, DC: The National Academies Press. doi: 10.17226/13975.
×
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Suggested Citation:"Chapter Two - Site and Geomaterial Characterization." National Academies of Sciences, Engineering, and Medicine. 2006. Rock-Socketed Shafts for Highway Structure Foundations. Washington, DC: The National Academies Press. doi: 10.17226/13975.
×
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Suggested Citation:"Chapter Two - Site and Geomaterial Characterization." National Academies of Sciences, Engineering, and Medicine. 2006. Rock-Socketed Shafts for Highway Structure Foundations. Washington, DC: The National Academies Press. doi: 10.17226/13975.
×
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Suggested Citation:"Chapter Two - Site and Geomaterial Characterization." National Academies of Sciences, Engineering, and Medicine. 2006. Rock-Socketed Shafts for Highway Structure Foundations. Washington, DC: The National Academies Press. doi: 10.17226/13975.
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SCOPE This chapter describes site investigation methods, classifica- tion systems for intact rock and rock masses, and field and laboratory tests used to determine rock engineering proper- ties. The focus is limited to information relevant to the design and construction of rock-socketed drilled shafts. Several references are available that provide guidance on strategies and methods of site characterization and material property evaluation for geotechnical practice, with a focus on trans- portation facilities. These include the FHWA Manual on Subsurface Investigations (Mayne et al. 2001), “Evaluation of Soil and Rock Properties,” Geotechnical Engineering Circular No. 5 (Sabatini et al. 2002), and the AASHTO Man- ual on Subsurface Investigations (1988). In addition, the U.S. Army Corps of Engineers has published several manuals rel- evant to this topic (Rock Testing Handbook 1993; Rock Foundations 1994; “Geotechnical Investigations” 2001). The purpose of site characterization is to obtain the infor- mation required to develop a model of the site geology and to establish the required engineering properties of the geomateri- als. The information obtained is used for two general purposes: (1) analysis of capacity and load-deformation response, which determines the foundation overall design; and (2) construction feasibility, costs, and planning. Once the site for a bridge or other transportation structure has been established, all aspects of the site and material characterization program are focused on the soil and rock conditions as they exist at that site. Geologic conditions and rock mass characteristics can exhibit such a wide degree of variability that it is not possible to estab- lish a single standardized approach. The scope of the program is determined by the level of complexity of the site geology, foundation loading characteristics, size, configuration, and structural performance of the bridge, acceptable levels of risk, experience of the agency, and other factors. Some of the infor- mation needed to establish the scope of site characterization may only be known following a preliminary study of the site. Rock and IGM exhibit behaviors that are unique and require special techniques for application to engineering problems. Two aspects of rock behavior that are paramount are: (1) natural rock masses may exhibit a high degree of variability and (2) properties of a rock mass are determined by the combined properties of intact rock and naturally occurring discontinuities, such as joints, bedding planes, faults, and other structural features. 8 SITE GEOLOGY Understanding the geologic environment provides informa- tion used to plan the more detailed, subsequent phases of ex- ploration. Site geology refers to the physiography, surficial geology, and bedrock geology of the site. The starting point is a thorough survey of existing information. In many cases, existing data will enable identification of geologic features that will determine the feasibility of rock-socketed founda- tions or will have a major impact on their design or con- struction. The amount and quality of information gathered can then be used to establish the type and extent of additional data that will be required. General knowledge of the site geology is required in the first phase of the design process outlined in chapter one, Conceptual Bridge Foundation Design, to establish anticipated site conditions, feasibility of rock sockets, and conceptual evaluation of potential geo- technical hazards. Sources of existing data include: geologic and topographic maps, publications, computer databases, aerial photographs, and consultation with other professionals. Many references are available that provide detailed information on sources and applications of existing data to geotechnical site characteriza- tion (e.g., Mayne et al. 2001). A detailed treatment of the topic is beyond the scope of this report and only the general aspects of such data sources will be summarized. Geologic maps are used to transmit information about geo- logic features at or near the earth’s surface. Maps are pre- pared at various scales and for a variety of purposes (Varnes 1974). A geologic map may be prepared to depict the general geology of a large region, for example bedrock geology of an entire state, or it may cover a relatively small area and con- tain detailed information about specific geologic features, for example engineering geology of a single quadrangle. A good starting point is the geologic map of the state. These maps are produced at a scale that makes it possible to identify the underlying bedrock formations in a general area. Often this is sufficient to know immediately whether a bridge is located where bedrock conditions are favorable or unfavorable for foundations in rock, or even whether bedrock exists at rea- sonable depth. Most state DOT geotechnical engineers and geologists with experience have familiarity with the geology of their state and incorporate this step unconsciously. The next logical step is to determine if more detailed geologic maps or reports are available for the particular area in which CHAPTER TWO SITE AND GEOMATERIAL CHARACTERIZATION

9the bridge is located. Sources of such maps and publications include U.S. Geological Survey and state Geological Sur- veys, university libraries, and Soil Conservation Service. The use of Internet search engines has added a powerful tool for locating such information and most governmental geologic publications can now be identified and obtained on-line. Detailed geologic maps normally provide useful information on characteristics of bedrock and, in some cases surficial, geology relevant to foundation engineering. These maps pro- vide descriptions of rocks in terms of lithology (rock type, mineralogy, and genesis), age, and structure (strike and dip of sedimentary rocks). In addition, major structural features are identified, such as faults, folds, and contacts between rock units (formations or members). Geologic maps prepared specifically for engineering purposes may include data on discontinuity patterns and characteristics, rock material strength, Rock Mass Ratings (RMRs), groundwater condi- tions, and depth to bedrock (Radbruch-Hall et al. 1987). Many will identify geologic hazards such as swelling soils or rock, landslides, corrosion potential, karst, abandoned mines, and other information of value. If engineering geo- logic maps are available they are an essential tool that should be used. The most practical aerial photographs for geotechnical purposes are black and white photographs taken with stereo overlap and with panchromatic film, from heights of between 500 m and 3,000 m, at scales of about 1:10,000 to 1:30,000. The higher level photographs provide a resolution most use- ful for larger-scale features such as topography, geology, and landform analysis, whereas the lower-level photographs pro- vide more detail on geologic structure. Landslides and debris flows, major faults, bedding planes, continuous joint sets, rock outcrops, and surface water are some of the features that can be identified and are relevant to the siting of bridge structures. A potentially valuable source of existing data may be consultation with other geoprofessionals with design or con- struction experience in the same rock units. Geotechnical engineers, geologists, groundwater hydrologists, contrac- tors, mining company personnel, well drillers, etc., may be able to provide geotechnical engineering reports from nearby projects, photographic documentation of excavations or other construction works, and unpublished reports or test- ing data. In addition, such individuals are often willing to share relevant experience. Bedian (2004) describes a case history in which experience at an adjacent site was used to develop a value engineer proposal for the design of rock- socketed foundations for a high rise building. The geotechnical literature contains many useful papers describing design, construction, and/or load testing of rock- socketed drilled shafts in which the focus is on a particular type of rock or a specific formation. For example, Hassan and O’Neill (1997) present correlations for side resistance of shafts in the Eagle Ford Shale, a rock unit commonly encountered in north-central Texas, most notably in the Dallas area. Results of load tests on drilled shafts in mica schist of the Wissahickon Formation, commonly encoun- tered in Philadelphia and other parts of eastern Pennsylvania, are given in Koutsoftas (1981) and Yang et al. (2004). Turner et al. (1993) and Abu-Hejleh et al. (2003) consider side re- sistance from load tests on shafts socketed into Pierre and Denver Formation shales. McVay et al. (1992) present a thor- ough study on the design of shafts in Florida limestone. Numerous other examples could be cited. Whenever such publications are available they should be used as a source of background information during the planning phase of any project where the same rock units are present. Results of load tests at different locations, but in the same rock unit, cannot be applied without judgment and site-specific considera- tions, but they do provide a framework for considering design issues and may provide insight on expected perfor- mance. Similarly, publications describing construction chal- lenges in certain geologic environments and strategies for addressing them can be useful. Schwartz (1987) described construction problems and recommended solutions for rock- socketed piers in Piedmont formations in the Atlanta area. Brown (1990) identified problems involved in construction of drilled shafts in the karstic limestone of northern Alabama and suggests methods and approaches that have been suc- cessful for dealing with such challenges. A literature review often is all that is necessary to locate this type of useful information. Where bedrock is exposed in surface outcrops or exca- vations, field mapping is an essential step to obtaining in- formation about rock mass characteristics relevant to design and construction of foundations. A site visit is recom- mended for reconnaissance and field mapping following a review of existing information. A competent engineering geologist or geotechnical engineer can make and record ob- servations and measurements on rock exposures that may complement, or in some cases exceed, the information ob- tained from borings and core sampling. Rock type, hard- ness, composition, degree of weathering, orientation and characteristics of discontinuities, and other features of a rock mass may be readily assessed in outcrops or road cuts. Guidance on detailed geologic mapping of rock for en- gineering purposes is given in Murphy (1985), Rock Slopes . . . (1989), and ASTM D4879 (Annual Book . . . 2000). Photography of the rock mass can aid engineers and contractors in evaluating potential problems associated with a particular rock unit. The major limitation lies in whether the surface exposure is representative of the rock mass at a depth corresponding to foundation support. When rock cor- ing and surface mapping demonstrate that surface exposures are representative, the surface exposures should be ex- ploited for information. Figure 4 shows a bridge site where mapping of rock exposures could provide much of the rele- vant data for design of foundations.

FIELD INVESTIGATIONS Field methods for characterization of rock include geophys- ical methods, rock core drilling, and in situ testing. These activities normally are carried out during the Preliminary Foundation Design phase of the design process as described in chapter one, and would be used to provide a description of subsurface conditions and a preliminary subsurface pro- file. The detailed results of field investigations, including detailed boring logs, in situ testing results, and interpreta- tion, would be included in the final geotechnical report pre- pared during the Final Foundation Design phase of Figure 2. Geophysical Methods Geophysical methods, in conjunction with borings, can pro- vide useful information in areas underlain by rock. The most common application of geophysics is to determine depth to bedrock. When correlated with data from borings, geophys- ical methods provide depth to bedrock information over a large area, eliminating some of the uncertainty associated with interpolations of bedrock depths for locations between borings. Geophysical methods are based on measuring the trans- mission of electromagnetic or mechanical waves through the ground. Signal transmission is affected by differences in the physical properties of geomaterials. By transmitting electro- magnetic or seismic signals and measuring their arrival at other locations, changes in material properties can be located. In some cases, the material properties can also be quantified. For foundation site characterization, geophysical methods can be placed into two general categories, those conducted from the ground surface (noninvasive) and those conducted in boreholes (invasive). When grouped according to method, the six major categories are: seismic, electromagnetic, electrical, magnetic, radar, and gravity. Basic descriptions of geophysi- cal methods and their application to geotechnical engineering are given by the U.S. Army Corps of Engineers (“Geophysi- cal Exploration . . . .” 1995) and Mayne et al. (2001). 10 NCHRP Synthesis 357: Use of Geophysics for Trans- portation Projects (Sirles 2006) provides a comprehensive overview of the topic and additional survey data relevant to this study. Table 1 identifies the primary and secondary methods used to investigate selected subsurface objectives. The table is an abridged version from the Sirles report (2006) in which only objectives pertaining to foundation investiga- tions are included. The survey of transportation agencies for this project identified “seismic” as the most widely used geophysical methods and “mapping rock” as the most widely used application of geophysics. Mapping karst or other voids was also identified as a major objective. Results of the survey for this study are consistent with those of Sirles (2006). The most frequently applied method is seismic refraction, which is based on measuring the travel time of compressional waves through the subsurface. Upon striking a boundary between two media of different proper- ties the direction of travel is changed (refraction). This change in direction is used to deduce the subsurface profile. Figure 5a illustrates the basic idea for a simple two-layer profile in which soil of lower seismic velocity (Vp1) overlies rock of higher seismic velocity (Vp2). A plot of distance from the source versus travel time (Figure 5b) exhibits a clear change in slope corresponding to the depth of the interface. The equipment consists of a shock wave source (typically a hammer striking a steel plate), a series of geophones to mea- sure seismic wave arrival, and a seismograph with oscillo- scope. The seismograph records the impact and geophone signals in a timed sequence and stores the data digitally. The technique is rapid, accurate, and relatively economical when applied correctly. The interpretation theory is relatively straightforward and equipment is readily available. The most significant limitations are that it is incapable of detecting material of lower velocity (lower density) underlying higher velocity (higher density) and that thin layers sometimes are not detectable. For these reasons, it is important not to rely exclusively on seismic refraction, but to verify depth to rock in several borings and correlate the seismic refraction signals to the boring results. Seismic velocity, as determined from seismic refraction measurements, can be correlated to small- strain dynamic modulus of soil and rock by the following relationships: (1) (2) in which Ed = small-strain dynamic modulus, vd = small- strain dynamic Poisson’s ratio, ρ = mass density, Vs = shear wave velocity, and Vp = compressional wave velocity. Eqs. 1 and 2 are based on the assumption that the rock mass is a homogeneous, isotropic, elastic solid. Because most rock masses depart significantly from this assumption, elastic modulus values calculated from seismic wave velocities are normally larger than values measured in static field load E v v v Vd d d p= −( ) +( ) −( ) 1 2 1 1 2ρ E v Vd d s= +( )2 1 2ρ FIGURE 4 Bridge site with surface exposures of foundation rock.

11 tests, such as plate bearing or pressure chamber tests. Alter- natively, a method that correlates rock mass modulus to shear wave frequency has been shown to provide a reasonable first- order estimate of modulus. Figure 6 shows the relationship between in situ modulus and shear wave frequency using a hammer seismograph, as described by Bieniawski (1978). The data can be fit to a straight line by EM = 0.054f – 9.2 (3) where EM = rock mass static modulus (GPa) and f = shear wave frequency (hertz) from the hammer blow received at distances of up to 30 m on a rock surface. Resistivity is a fundamental electrical property of geo- materials that varies with material type and water content. To measure resistivity from the ground surface (Figure 7), elec- trical current is induced through two current electrodes (C1 and C2), while change in voltage is measured by two poten- tial electrodes (P1 and P2). Apparent electrical resistivity is then calculated as a function of the measured voltage differ- ence, the induced current, and spacing between electrodes. Two techniques are used. In a sounding survey, the center- line of the electrodes is fixed while the spacing of the elec- trodes is increased for successive measurements. The depth of material subjected to current increases with increasing electrode spacing. Therefore, changes in measured apparent resistivity with increasing electrode spacing are indicative of a change in material at depth. In this way, variations in material properties with depth (layering) can be determined. The second method is a profiling survey in which the elec- trode spacing is fixed but the electrode group is moved horizontally along a line (profile) between measurements. Changes in measured apparent resistivity are used to deduce lateral variations in material type. Electrical resistivity methods are inexpensive and best used to complement seis- mic refraction surveys and borings. The technique has ad- vantages for identifying soft materials in between borings. Limitations are that lateral changes in apparent resistivity can be interpreted incorrectly as depth related. For this and other reasons, depth determinations can be in error, which is why it is important to use resistivity surveys in conjunc- tion with other methods. The use of multi-electrode resistivity arrays shows promise for detecting detailed subsurface profiles in karst terranes, one of the most difficult geologic environments for rock-socketed foundations. Dunscomb and Rehwoldt (1999) showed that two-dimensional (2-D) profiling using multi-electrode arrays provides reasonable resolution for imaging features such as pinnacled bedrock surfaces, overhanging rock ledges, frac- ture zones, and voids within the rock mass and in the soil overburden. Hiltunen and Roth (2004) present the results of multiple-electrode resistivity surveys at two bridge sites on I-99 in Pennsylvania. The resistivity profiles were com- pared with data from geotechnical borings. Both sites are located in karst underlain by either dolomite or limestone. The resistivity profiles provided a very good match to the Techniques Methods Investigation Objectives Bedrock Depth P P P P P P S S Rippability P P P P S S S S S S S S P PS SP P P P P PP Lateral & Vertical Variation in Rock or Soil Strength Location of Faults and Fracture Zones Karst Features M ag ne tic s G ro un d Pe ne tra tin g Ra da r Ti m e– D om ai n EM S ou nd in gs El ec tri ca l R es ist iv ity /P El ec tri ca l R es ist iv ity T o m o gr ap hy /P G ra vi ty Se ism ic R ef ra ct io n Se ism ic R ef le ct io n Se ism ic T o m o gr ap hy Sh ea r W av e Su rfa ce W av e (S AS W, M A SW , & Pa ss iv e) EM 31 — Te rr ai n Co nd uc tiv ity EM 34 — Te rr ai n Co nd uc tiv ity EM 61 — Ti m e– D om ai n M et al D et ec to r Seismic Electromagnetic Electrical Other Notes: P = primary; S = secondary; blank = techniques should not be used; EM = electromagnetic; SASW = spectral analysis of surface waves; MASW = multi-channel analysis of surface waves. TABLE 1 GEOPHYSICAL METHODS AND APPLICATIONS (after Sirles 2006)

stratigraphy observed in borings, particularly for top-of- rock profile. Figure 8 shows a resistivity tomogram at one of the bridge pier sites, in which the top-of-rock profile is well-defined by the dark layer. Inclusions of rock in the overlying soil are also clearly defined. This technology should be considered for any site where a rock surface pro- file is required and would provide valuable information for both design and construction of rock-socketed founda- tions. Table 1 identifies electrical resistivity tomography profiling as a primary method for investigating karstic con- ditions and as a secondary method for measuring depth to bedrock. Other geophysical methods have potential for rock sites, but have yet to be exploited specifically for applications to foundations in rock. These include downhole and crosshole seismic methods. Downhole seismic p-wave is based on mea- suring arrival times in boreholes of seismic waves generated 12 at the ground surface. Crosshole seismic involves measur- ing travel times of seismic waves between boreholes. Both methods provide depth to rock, and s-wave velocities, dynamic shear modulus, small-strain Young’s modulus, and Poisson’s ratio. Crosshole tomography is based on computer analysis of crosshole seismic or resistivity data to produce a 3-dimensional (3-D) representation of subsurface conditions. These techniques are more expensive and require specialized expertise for data interpretation, but may be cost-effective for large structures where the detailed information enables a more cost-effective design or eliminates uncertainty that may otherwise lead to construction cost overruns. All geophysical methods have limitations associated with the underlying physics, the equipment, and the individuals running the test and providing interpretation of the data. The study by Sirles (2006) includes several informative case his- tories from state DOTs of both successful and unsuccessful projects. The single case history related to a bridge founda- tion investigation is one of a failure to provide accurate (a) (b) FIGURE 5 Seismic refraction method (Mayne et al. 2001): (a) field setup and procedures; (b) data reduction for depth to hard layer. FIGURE 6 Rock mass modulus versus shear wave frequency by Bieniawski (Goodman 1980). C2 C1 P2 P1 Spacing, A Spacing, A V Spacing, A Battery Current meter Volt meter FIGURE 7 Field configuration for resistivity test.

13 depths to bedrock in a river channel using both seismic refraction and an electrical resistivity sounding survey. Rea- sons cited for the failure include loss of geophones owing to running water and ice, instrumentation malfunctions, ex- cessive background noise, differences of opinion between consultants on data interpretation, and discrepancies between top of rock from geophysical results and borings. Although this is not believed to be a typical case, it demonstrates some real world lessons. Additional findings by Sirles (2006) are that “in-house geoscientists and engineers do not understand the value, the benefit, or the science of geophysics for their projects.” How- ever, several factors point to geophysics becoming more widely accepted and implemented as a tool in the transporta- tion industry. These include a manual published by FHWA and available on-line (http://www.cflhd.gov/geotechnical), additional programs aimed at training of agency personnel, and increasing levels of experience. Borings Borings provide the most direct evidence of subsurface con- ditions at a specific site. They furnish detailed information on stratigraphy and samples of soil and rock from which engineering properties are determined. Borings also provide the means for conducting in situ tests, installation of instru- mentation, and observing groundwater conditions. Conven- tional soil boring and testing equipment is used to drill through overlying soil deposits and to determine depth to bedrock. Once encountered, the most widely used technique for investigating rock for the purpose of foundation design is core drilling. Samples are obtained for rock classification and determining rock properties important to both design and construction. A core sample can be examined physically and tested, providing information that is hard to obtain by any other methods. Rock core drilling is accomplished using rotary drill equipment, usually the same truck- or skid-mounted rigs used for soil drilling and sampling. A hollow coring tube equipped with a diamond or tungsten–carbide cutting bit is rotated and forced downward to form an annular ring while preserving a central rock core. Standard core barrel lengths are 1.5 m and 3 m (5 ft and 10 ft). Fluid, usually water but possibly drilling mud, is circulated for cooling at the cutting interface and removal of cuttings. Selecting the proper tools and equipment to match the conditions and the expertise of an experienced drill crew are essential elements of a suc- cessful core drilling operation. Once rock is encountered, coring normally is continuous to the bottom of the hole. Where the rock being sampled is deep, wire line drilling, in which the core barrel is retrieved through the drill stem, eliminates the need to remove and reinsert the entire drill stem and can save considerable time. If sampling is not con- tinuous, drilling in between core samples can be accom- plished using solid bits. Rock coring bits and barrels are available in standardized sizes and notations. Important considerations in core barrel selection are: (1) core recovery and (2) the ability to deter- mine the orientation of rock mass structural features relative to the core. Core recovery is most important in highly frac- tured and weak rock layers, because these zones are typically critical for evaluation of foundation–rock load transfer. For sampling of competent rock, bits and core barrels that provide a minimum of 50-mm-diameter (nominal) core are adequate for providing samples required for index tests, rock quality designation (RQD), laboratory specimens for strength testing, and evaluating the conditions of discontinu- ities. For example, NWM (formerly NX) diamond bit and rock core equipment drills a 76-mm (3-in.) diameter hole and provides a 54-mm (2.125-in.) diameter rock core. When weak, soft, or highly fractured rock is present, it may be necessary to use larger diameter bits and core barrels to improve core recovery and to obtain samples from which laboratory strength specimens can be prepared. Coring tools up to 150 mm (6 in.) in diameter are used. A highly recommended practice for best core recovery is to use triple-tube core barrels. The inner sampling tube does not rotate during drilling and is removed by pushing instead of hammering; features that minimize disturbance. Thorough descriptions of coring equipment and techniques are given in Acker (1974), AASHTO Manual on Subsurface Investigations (1988), Mayne et al. (2001), and U.S. Army Corps of Engineers (“Geotechnical Investigations” 2001). Steeply dipping or near-vertical bedding or jointing may go undetected in holes drilled vertically (Terzaghi 1965). Such features can significantly influence the strength and de- formability of rock foundations. Inclined (nonvertical) drilling provides the opportunity to detect the orientation and characteristics of near-vertical features. Oriented core refers to any method that provides a way to determine the geomet- rical orientation of planar structural features, such as bed- ding, joints, fractures, etc., with respect to the geometrical orientation of the core. One approach is to mark the core with a special engraving tool so that the orientation of the discon- tinuity relative to the core is preserved and the orientation of 10 20 30 40 50 60 70 80 90 100 -20 -10 0 12 25 50 100 200 Resistivity (Ohm-feet) Distance (feet) D ep th (fe et) East West Resistivity Test #7 FIGURE 8 Resistivity tomogram at Pennsylvania bridge site in karst (Hiltunen and Roth 2004).

the discontinuity (strike and dip) can be determined accu- rately (Goodman 1976). A method used with wire line drilling involves making an impression of the core in clay. The combination of inclined and oriented coring techniques can provide an effective tool for characterizing orientation of discontinuities in complexly fractured rock masses. Rock core orienting methods are covered in more detail in the AASHTO Manual on Subsurface Investigations (1988) and are also reviewed and compared with borehole televiewer methods by Eliassen et al. (2005). Depth and Spacing of Boreholes O’Neill and Reese (1999) recommend the number of borings to be made per drilled shaft location at bridge sites when the material to be excavated is unclassified (Table 2). Unclassi- fied means the contractor is paid by the unit of excavation depth (meters or feet) regardless of the material encountered. For rock sites, these recommendations should be considered a minimum. If possible, it is recommended to locate one boring at every rock-socketed shaft. In practice, this is not always possible and factors such as experience, site access, degree of subsurface variability, geology, and importance of the structure will be considered. If materials are classi- fied for payment purposes, it becomes more important to locate a boring at every drilled shaft location for the purpose of making accurate cost estimates and for contractors to base their bids on knowledge of the materials to be exca- vated. Where subsurface conditions exhibit extreme varia- tions over short distances, multiple borings at each shaft location can reduce the risk of founding a shaft on soil in- stead of rock. For example, large-diameter, nonredundant shafts in karstic limestone may require multiple borings at each shaft location to determine that the entire base will be founded in rock and to identify voids or zones of soil beneath the base that may affect load-settlement behavior of each shaft. The draft 2006 Interim AASHTO LRFD Bridge Design Specifications recommends the following for depth of borings below anticipated tip elevations: 14 For shafts supported on or extending into rock, a minimum of 3 m of rock core, or a length of rock core equal to at least three times the shaft diameter for isolated shafts or two times the max- imum shaft group dimension, whichever is greater, shall be extended below the anticipated shaft tip elevation to determine the physical characteristics of rock within the zone of foundation influence. If the tip elevation changes at some point during the project, additional drilling may be required to meet this recommenda- tion. O’Neill and Reese (1999) provide the following guidance on boring depth. When the RQD is less than 50%, extend bor- ing depths to at least 125% of the expected depths of the drilled shaft bases plus two base diameters. If RQD values are greater than approximately 50% at the planned base elevation, borings only need be extended to the expected base elevation plus two base diameters as long as the RQD remains above 50%. The rationale is that it is not likely the shafts will need to be deep- ened once the actual strata are exposed. This approach requires that foundation diameters and depths be estimated before the boring program and that RQD be determined during drilling. The approach described is only a general suggestion and local geologic conditions may dictate other criteria for boring depths. If in the course of design or construction it becomes necessary to deepen the shafts, supplementary borings should be taken. An available, but not widely used tool for subsurface inves- tigation is to drill one or more large-diameter borings or to have a drilled shaft contractor install a full-sized test excavation. Large-diameter borings can be made with augers in soft rock and with core barrels in hard rock. The sidewalls of the boring or shaft can be examined directly (with appropriate safety mea- sures) or with downhole cameras. Observations can then be made of rock mass features, including degree of roughness and general quality of the drilled surfaces, and fracture patterns. Large-diameter holes provide access for obtaining high-quality undisturbed samples and may be used for performing in situ plate load tests to measure rock mass modulus. If a full-size excavation is made by a drilled shaft contractor, information of value to both engineers and contractors is obtained. In Fig- ure 3, “constructability” is one of the items to be determined during the site characterization. A full-sized excavation is the most direct method for obtaining this information. Downhole devices are available for borehole viewing and photography, including borescopes, photographic cameras, and television cameras. A visual image of rock in the sidewalls of a boring provides information on structural features that may add significantly to the overall picture of subsurface geology. Advantages and disadvantages of some remote viewing devices are discussed in “Geotechnical Investigations” (2001); however, the technologies for borehole imaging are advancing rapidly and the user should consult commercial providers for the most up-to-date information. These devices are effective for examining soft zones for which core may not have been recovered, determination of dip and strike of important struc- tural features, and viewing of cavities such as solution voids, open joints, and lava tunnels in volcanic rocks. Redundancy Condition Shaft Diameter (m) Guideline Single-column, single shaft foundations All One boring per shaft Redundant, multiple- shaft foundations >1.8 m (6 ft) One boring per shaft Redundant, multiple- shaft foundations 1.2–1.8 m (4–6 ft) One boring per two shafts Redundant, multiple- shaft foundations <1.2 m (4 ft) One boring per four shafts Source: OíNeill and Reese 1999. TABLE 2 RECOMMENDED FREQUENCY OF BORINGS, DRILLED SHAFT FOUNDATIONS FOR BRIDGES, UNCLASSIFIED EXCAVATION

15 Borehole televiewers provide high-resolution images showing rock mass structural and textural features and ac- curate measurement of dip and dip direction of structural features without the use of oriented core. Optical teleview- ers (OTV) generate a high-resolution digital color image of the inside of the borehole wall and are capable of resolving fractures as narrow as 0.1 mm with a radial resolution of 1 degree (Eliassen et al. 2005). The OTV can be operated in air- or fluid-filled boreholes; however, fluid requires thor- ough flushing before image acquisition is undertaken. Acoustic televiewers (ATV) produce images of the borehole wall based on the amplitude and travel time of acoustic signals reflected from the borehole wall. A portion of the reflected energy is lost in voids or fractures, producing dark bands on the amplitude log. Travel time measurements allow recon- struction of the borehole shape, making it possible to generate a 3-D representation of a borehole. Both types of televiewers orient their image data using a three-component fluxgate magnetometer and a three- component tilt meter incorporated into the tool. Before inter- pretation, the image is rotated to a common reference direction, either magnetic north or the high side of the borehole. Planar features that intersect the borehole wall produce sinusoidal traces in the “unwrapped,” or 2-D, televiewer image. Using the reference direction recorded during logging, sinusoids can be analyzed to produce dip and dip directions of structural features. Figure 9 shows OTV and ATV images of the same borehole and illustrates some advantages of each device. The OTV is able to provide a color image of the dike and excel- lent imaging of the texture of the granite. The ATV highlights fracturing within the diorite. The California DOT (Caltrans) reports using the ATV to provide very-high-resolution sonic images in the format of a 3-D “pseudo-core,” as illustrated in Figure 10. According to Eliassen et al. (2005), use of optical and acoustic televiewer equipment is gaining popularity over oriented coring techniques because it is generally less labor intensive and is particularly useful where access or ability to drill inclined holes is limited or where local drilling compa- nies lack the equipment necessary to collect oriented cores. However, to date, this technology is being applied to site char- acterization for rock slope engineering and underground openings, and is not being used in foundation investigations. Eliassen et al. (2005) note further that televiewer logs are best used to supplement data obtained from quality rock coring, which provides samples for laboratory testing, assessment of joint and discontinuity planes, and correlation of lithologic and geologic boundaries with geophysical data. The authors suggest that drilling time and costs can be optimized with ap- propriate combinations of coring and less expensive air rotary boreholes logged with OTV and ATV equipment. Borehole televiewing may be most useful in rock-socket applications at sites where the structural orientation of discontinuities is a significant factor in foundation stability. For example, some modes of bearing capacity failure (described in chapter three) depend on the orientation of discontinuities in the rock mass below the socket base. LaFronz et al. (2003) describe use of OTV as part of the subsurface investigation for the Colorado River Bridge at Hoover Dam. The primary purpose was to obtain structural data to develop recommendations for exca- vation of cut slopes at the abutment foundations. FIGURE 9 Optical and acoustic televiewer images of a 50-cm diorite dike in granite (Eliassen et al. 2005). FIGURE 10 An acoustic television log (Caltrans 2005).

GEOLOGIC AND INDEX PROPERTIES OF ROCK The most basic characterization of rock for engineering purposes is a description of rock core based on visual and physical examination. The International Society of Rock Mechanics (ISRM) proposed a standardized method for descriptions of rock masses from mapping and core logging (“Basic Geotechnical Description of Rock Masses” 1981). A summary of the ISRM method as given by Wyllie (1999) is adopted in the FHWA manuals on subsurface investigations and soil and rock properties (Mayne et al. 2001; Sabatini et al. 2002) and is summarized here. A rock mass is described in terms of five categories of properties, as follows: 1. Rock Material Description—a. Rock type, b. Wall strength, c. Weathering 2. Discontinuity Description—d. Type, e. Orientation, f. Roughness, g. Aperture 3. Infilling—h. Infilling type and width 4. Rock Mass Description—i. Spacing, j. Persistence, k. Number of sets, l. Block size/shape 5. Groundwater—m. Seepage. Each of the 13 parameters listed (a through m) is assigned a description using standardized terminology. Descriptive terms are given in Tables 3 through 6 and in Figure 11, which is an example of a Key used for entering rock descriptions on a coring log and includes details of several categories. Rock Material Descriptors Rock type is defined in terms of origin (igneous, sedimentary, or metamorphic) and then further classified into one of the 16 rock types listed in Table 3 based on lithologic characteristics that include color, fabric (microstructural and textural fea- tures), grain size and shape (Tables 4 and 5), and mineralogy. Sedimentary rock descriptions should include bedding thick- ness (Table 6). The rock unit name, which may be a formal name of a formation or an informal local name, should be identified; for example, Bearpaw Shale or Sherman Granite. Compressive strength of rock core can be evaluated us- ing simple field tests with equipment commonly available (knife, rock hammer, etc.) and summarized in the Key of Figure 11 (“Rock Strength”) or evaluated from point load Igneous Intrusive (coarse-grained) Extrusive (fine-grained) Pyroclastic Granite Syenite Diorite Diabase Gabbro Peridotite Pegmatite Rhyolite Trachyte Andesite Basalt Obsidian Pumice Tuff Sedimentary Clastic (sediment) (chemically formed) (organic remains) Shale Mudstone Claystone Siltstone Conglomerate Limestone, oolitic Limestone Dolomite Gypsum Halite Chalk Coquina Lignite Coal Metamorphic Foliated Nonfoliated Slate Phyllite Schist Gneiss Quartzite Amphibolite Marble Hornfels Description Diameter (mm) Characteristic Very coarse grained Coarse grained Medium grained Fine grained Very fine grained >4.75 2.00–4.75 0.425–2.00 0.075–0.425 <0.075 Grain sizes are greater than popcorn kernels Individual grains can be easily distinguished by eye Individual grains can be distinguished by eye Individual grains can be distinguished with difficulty Individual grains cannot be distinguished by unaided eye TABLE 4 TERMS TO DESCRIBE GRAIN SIZE OF SEDIMENTARY ROCK TABLE 5 TERMS TO DESCRIBE GRAIN SHAPE (for sedimentary rocks) Description Characteristic Angular Subangular Subrounded Rounded Well-rounded Showing very little evidence of wear. Grain edges and corners are sharp. Secondary corners are numerous and sharp. Showing definite effects of wear. Grain edges and corners are slightly rounded off. Secondary corners are slightly less numerous and slightly less sharp than in angular grains. Showing considerable wear. Grain edges and corners are rounded to smooth curves. Secondary corners are reduced greatly in number and highly rounded. Showing extreme wear. Grain edges and corners are smoothed off to broad curves. Secondary corners are few in number and rounded. Completely worn. Grain edges and corners are not present. No secondary edges or corners are present. TABLE 3 ROCK GROUPS AND TYPES

17 tests or uniaxial compression tests conducted on specimens. The rock strength descriptions given at the bottom of the second page of the Key correspond to the seven categories of rock strength, R0 through R6, of the ISRM (“Basic Geo- technical Description of Rock Masses” 1981), with R0 cor- responding to extremely weak rock and R6 corresponding to extremely strong rock. The degree of physical disinte- gration or chemical alteration of rock can be described by the terms and abbreviations given in the Key. Weathering and alteration reduces shear strength of both intact rock and discontinuities. TABLE 6 TERMS TO DESCRIBE STRATUM THICKNESS Descriptive Term Stratum Thickness Very thickly bedded Thickly bedded Thinly bedded Very thinly bedded Laminated Thinly laminated >1 m 0.5 to 1.0 m 50 mm to 500 mm 10 mm to 50 mm 2.5 mm to 10 mm <2.5 mm FIGURE 11 Key for rock core description (sheet 1).

Discontinuity Descriptors A discontinuity is defined as any surface across which any me- chanical property of a rock mass is discontinuous. Discon- tinuity descriptors are summarized in Figure 11 (Key), items a through g. Types of discontinuities include faults, joints, shear planes, foliation, veins, and bedding. Orientation refers to the measured dip and dip direction of the surface (or dip and strike). Dip is defined as the maximum angle of the plane to the horizontal and dip direction (strike) is the direction of the horizontal trace of the line of dip measured clockwise from north, in degrees. Determination of dip and dip direction from core samples is possible using oriented coring techniques, borehole televiewers, downhole cameras, or other devices capable of establishing orientation of the discontinuity relative 18 to the core. Roughness and surface shape of joint surfaces is best measured in the field on exposed surfaces at least 2 m in length and can be described using the terms in the Key or quantified in terms of a Joint Roughness Coefficient (Barton 1973). Aperture is the width of a discontinuity with no infill- ing and can be classified according to Box c of the Key. Infilling Infilling is the term for material separating adjacent rock walls of discontinuities. Infilling is described in terms of its type, amount, and width (Key). Additional laboratory testing may be conducted to determine soil classification and shear strength of infilling materials. Direct shear tests provide a FIGURE 11 (continued ) (sheet 2).

19 means to measure shear strength of joints with infilling, as described by Wyllie and Norrish (1996). Infilling properties vary widely and can have a significant influence on rock mass strength (RMS), compressibility, and permeability. Rock Mass Descriptors Spacing is the perpendicular distance between adjacent dis- continuities. Spacing has a major influence on seepage and mechanical behavior and can be described using the terms in Figure 11 (Key). Persistence refers to the continuous length or area of a discontinuity and requires field exposures for its determination. The number of sets of intersecting discontinuities has a major effect on RMS and compressibility. As the number of sets increases, the extent to which the rock mass can deform without failure of intact rock also increases. Field mapping or observations made in exploratory pits or large excavations provide the best opportunity to map multiple sets of discon- tinuities. Block size and shape is determined by spacing, per- sistence, and number of intersecting sets of discontinuities. Descriptive terms include blocky, tabular, shattered, and columnar, while size ranges from small (<0.0002 m3) to very large (>8 m3). Seepage Field observations of seepage from discontinuities should be described whenever it can be observed. The presence and type of infilling controls joint permeability and should be described wherever seepage is observed. Seepage can range from dry to continuous flow under high pore water pressure Rock Quality Designation A simple and widely used measure of rock mass quality is provided by the RQD (rock quality designation, ASTM D6032). RQD is equal to the sum of the lengths of sound pieces of core recovered, greater than 100 mm (4 in.) in length, expressed as a percentage of the length of the core run. Originally introduced by Deere (1964), the RQD was evaluated by Deere and Deere (1989), who recommended modifications to the original procedure after evaluating its field use. Figure 12 illustrates the recommended procedure. Several factors must be evaluated properly for RQD to pro- vide reliable results. RQD was originally recommended for NX size core, but can also be used with the somewhat smaller NQ wireline sizes and with larger wire line sizes and other core sizes up to 150 mm (6 in.). RQD based on the smaller BQ and BX cores or with single-tube core barrels is discouraged because of core breakage. Core segment lengths should be measured along the centerline or axis of the core, as shown in Figure 12. Only natural fractures such as joints or shear planes should be considered when calculating RQD. Core breaks caused by drilling or handling should be fitted together and the pieces counted as intact lengths. Drilling breaks may be identified by fresh surfaces. For some laminated rocks it may be difficult to distinguish natural fractures from those caused by drilling. For characterization of rock mass behavior rele- vant to foundation design it is conservative to not count the length near horizontal breaks. RQD should be performed as soon as possible after the core is retrieved to avoid the effects of deterioration, which may include slaking and separation of core along bedding planes, especially in moisture-sensitive rocks like some shales. It is also desirable because RQD is a quantitative measure of core quality at the time of drilling when the rock core is “fresh” and most representative of in situ conditions. Rock assigned a weathering classification of “highly weath- ered” or above should not be included in the determination of RQD. RQD measurements assume that core recovery is at or near 100%. As core recovery varies from 100%, explanatory notes may be required to describe the reason for the variation and the effect on RQD. In some cases, RQD will have to be determined on the basis of total length of core recovered, rather than on the length of rock cored. One state (Florida) uses per- cent core recovery as an index of rock quality in limestone. A general description of rock mass quality based on RQD is given here. Its wide use and ease of measurement make it an important piece of information to be gathered on all core holes. Taken alone, RQD should be considered only as an L = 250 mm L = 0 HIGHLY WEATHERED DOES NOT MEET SOUNDNESS REQUIREMENT L = 200 mm L = 190 mm L = 0 CENTER LINE PIECES < 4" & HIGHLY WEATHERED L = 0 < 4" L = 0 NO RECOVERY LENGTHRUNCORETOTAL mm100 PIECESCORE SOUNDOFLENGTH RQD > = ∑ 100% 1200 200190250RQD ×++= ( )FAIR53%RQD = m m 0021 = HT G NEL L AT OT N U R E R O C MECHANICAL BREAK CAUSED BY DRILLING PROCESS FIGURE 12 RQD determination of rock core (after Deere and Deere 1989).

approximate measure of overall rock quality. RQD is most useful when combined with other parameters accounting for rock strength, deformability, and discontinuity characteris- tics. As discussed in subsequent sections of this report, many of the rock mass classification systems in use today incorpo- rate RQD as a key parameter. Rock Mass Description RQD Excellent 90–100 Good 75–90 Fair 50–75 Poor 25–50 Very Poor <25 ENGINEERING PROPERTIES OF ROCK Laboratory Tests on Intact Rock Intact rock refers to the consolidated and cemented assem- blage of mineral particles forming the rock material, ex- cluding the effects of macro-scale discontinuities such as joints, bedding planes, minor faults, or other recurrent pla- nar fractures. The term rock mass is used to describe the sys- tem comprised of intact rock and discontinuities. The char- acteristics of intact rock are determined from hand specimens or rock core. Properties of intact rock required for proper characterization of the rock mass and that are rele- vant to foundation design include strength and deformabil- ity. For some rock types, the potential for degradation on ex- posure to atmospheric conditions may also need to be evaluated. Some design methods incorporate properties of intact rock directly; for example, correlations between ulti- mate unit side resistance and uniaxial compressive strength. However, most analytical treatments of foundation capacity and load-deformation response incorporate the strength and deformability of intact rock into rock mass models that also 20 account for the effects of discontinuities, rock quality, and other factors. Table 7 lists the laboratory tests for intact rock most com- monly done for foundation design and gives the ASTM Standard Designation for each test. More thorough coverage of laboratory testing of intact rock is given by Mayne et al. (2001), the Rock Testing Handbook (1993), and the AASHTO Manual on Subsurface Investigations (1988). Engineering properties of intact rock that are used most often for foundation design are uniaxial compressive strength (qu) and elastic modulus (ER). The compressive strength of intact rock is determined by applying a vertical compressive force to an unconfined cylindrical specimen prepared from rock core. The peak load is divided by the cross-sectional area of the specimen to obtain the uniaxial compressive strength (qu). The ASTM procedure (D2938) specifies tolerances on smoothness over the specimen length, flatness of the ends, the degree to which specimen ends are perpendicular to the length, and length-to-diameter ratio. Uniaxial compressive strength of intact rock is used in empirical correlations to evaluate ultimate side and base resistances under axial loading; ultimate limit pressure under lateral loading; and, by contractors, to as- sess constructability. Elastic modulus of intact rock is measured during conduct of the uniaxial compression test by measuring deformation as a function of load. It is common to measure both axial and di- ametral strain during compression to determine elastic mod- ulus and Poisson’s ratio. Test procedures are given in ASTM Standard (D3148) and discussed further by Wyllie (1999). It is important to note that the ASTM procedure defines several methods of determination of modulus, including tangent modulus at a specified stress level, average modulus over the Test Category Name of Test and ASTM Designation Comments Uniaxial compression Unconfined compressive strength of intact rock core specimen (D2938) Primary test for strength and deformability of intact rock; input parameter for rock mass classification systems Split tensile Splitting tensile strength of intact rock core specimens (D3967) Splitting tensile strength of a rock disk under a compression line load Point load strength Determination of the point load strength index of rock (D5731) Index test for rock strength classification; can be performed in field on core pieces unsuitable for lab testing Direct shear Laboratory direct shear strength tests for rock specimens under constant normal stress (D5607) Applies to intact rock strength or to shear strength along planes of discontinuities, including rock–concrete interface Strength- deformation Elastic moduli of intact rock core specimens in uniaxial compression (D3148) Young’s modulus from axial stress–strain curve; Poisson’s ratio can also be determined Durability Slake durability of shales and similar weak rocks (D4644) Index test to quantify the durability of weak rocks under wetting and drying cycles with abrasion TABLE 7 COMMON LABORATORY TESTS FOR INTACT ROCK

21 linear portion of the stress–strain curve, and secant modulus at a fixed percentage of maximum strength. For rocks that exhibit nonlinear stress–strain behavior, these methods may provide significantly different values of modulus and it is important to note which method was used when reporting val- ues of modulus. The point load test is conducted by compressing a core sample or irregular piece of rock between hardened steel cones (Figure 13), causing failure by the development of ten- sile cracks parallel to the axis of loading. The uncorrected point load strength index is given by Is = P/D2 (4) where P = load at rupture, and D is the distance between the point loads. The point load index is reported as the point load strength of a 50 mm core. For other specimen sizes a correc- tion factor is applied to determine the equivalent strength of a 50 mm specimen. The point load index is correlated to uni- axial compressive strength by qu = C Is(50) (5) where qu is the unconfined compressive strength, Is(50) is the point load strength corrected to a diameter of 50 mm, and C is a correlation factor that should be established on a site- specific basis by conducting a limited number of uniaxial compression tests on prepared core samples. If a site-specific value of C is not available, the ASTM Standard recommends approximate values based on core diameter. For a 54 mm core (NX core size), the recommended value of C is 24. The principal advantages of the point load test are that it can be carried out quickly and inexpensively in the field at the site of drilling and that tests can be conducted on irregular specimens without the preparation required for uniaxial com- pression tests. Split tensile strength (qt) of rock (ASTM D4644) is deter- mined by compressing a cylindrical disk under a compressive line load. Split tensile strength has been correlated with unit side resistance; for example, by McVay et al. (1992) for drilled shafts in Florida limestone. Direct shear testing is applicable to determination of the Mohr–Coulomb shear strength parameters cohesion, c, and friction angle, φ, of discontinuity surfaces in rock (ASTM D5607). Shear strength of discontinuities may govern capac- ity in certain conditions; for example, base capacity of sock- eted foundations when one or two intersecting joint sets are oriented at an intermediate angle to horizontal. The other no- table application of this test is in simulating the shear strength at the rock–concrete interface for evaluation of side resistance of socketed shafts under axial loading. However, for this application, the constant normal stiffness (CNS) di- rect shear test described by Johnston et al. (1987) is more ap- plicable. Instead of a constant normal load, normal force is applied through a spring that increases or decreases the ap- plied force in proportion to the magnitude of normal dis- placement (dilation). Dilatancy of the interface is a major factor controlling strength and stiffness of socketed shafts under axial load. The slake durability test (ASTM D4644) provides an index for identifying rocks that will weather and degrade rapidly. The test is appropriate for argillaceous sedimentary rocks (mudstone, shale, clay–shales) or any weak rock. Representa- tive rock fragments are placed in a wire mesh drum and dried in an oven to constant weight. The drum is partially sub- merged in water and rotated at 20 revolutions per minute for a period of 10 min. The drum and its contents are then dried a second time and the loss of weight is recorded. The test cycle is repeated a second time and the slake durability index, ID, is calculated as the ratio (reported as a percentage) of final to ini- tial dry weights of the sample. Rocks with ID < 60 are consid- ered prone to rapid degradation and may indicate a suscepti- bility to degradation of the borehole wall when water is introduced during drilling, potentially leading to formation of a “smear zone.” Hassan and O’Neill (1997) define the smear zone as a layer of soil-like material along the socket wall and demonstrate that smearing can have a significantly negative effect on side load transfer of shafts in argillaceous rock.FIGURE 13 Point load test setup.

In Situ Tests for Rock In situ testing can be used to evaluate rock mass deformation modulus and, in some instances, RMS. In situ testing meth- ods with potential applications to rock-socket design are presented in Table 8. In situ testing of rock is not performed routinely for rock-socket design by most of the agencies surveyed for this study. The survey responses indicate that five state DOTs currently use the pressuremeter test (PMT) to obtain design parameters. Of these, all five use the test to obtain rock mass modulus. One state reported the use of PMT to evaluate RMS in weak rocks. Four states use the PMT for correlating test results with the parameters that de- fine p-y curves for analysis of shafts under lateral loading (chapter four). The term dilatometer is also used to describe a pressuremeter intended for use in rock but should not be con- fused with the flat plate dilatometer used for in situ testing of soil. One state (Massachusetts) reported using the borehole jack to measure rock mass modulus. No states reported using the plate load test for rock-socket design. Information on conduct and interpretation of the tests identified in Table 8 and other in situ tests for rock are given in the relevant ASTM standards, Rock Testing Handbook (1993) and Mayne et al. (2001). Heuze (1980) investigated the effect of test scale on the modulus of rock masses. Several types of field tests, includ- ing borehole jack and plate load tests at different scales, were included and results were compared with those of laboratory compression tests. It was observed that in situ rock mass modulus values generally range from 20% to 60% of intact 22 rock modulus from laboratory uniaxial compression tests. The borehole jack was recommended as a field test that, with proper analysis (Heuze 1984), yields values of rock mass modulus that are consistent with results from large plate bear- ing tests. The borehole jack designed for NX sized borings (75 mm or 3 in. diameter) affects a “test volume” of approxi- mately 0.14 m3 (5 ft3). Borehole jack devices are available commercially with limit pressures of up to 69 MPa, allowing the test to reach stress levels beyond the elastic limit and, for some weak rock masses, to ultimate strength. Studies on the use of PMTs for determination of rock mass modulus include those of Rocha et al. (1970), Bukovansky (1970), Georgiadis and Michalopoulos (1986), and Littlechild et al. (2000). Results have been mixed, with some research- ers indicating a high degree of agreement between PMT modulus and other in situ tests (e.g., Rocha et al. 1970) and others reporting PMT modulus values significantly lower than modulus measured by plate-load and borehole jack tests (e.g., Bukovansky 1970). Littlechild et al. (2000) concluded that PMTs, using the Cambridge High Pressure Dilatometer, were not useful for determination of rock mass modulus for design of deep foundations in several rock types in Hong Kong. In strong and massive rocks such as metasiltstone and tuff, the device did not have sufficient capacity to measure modulus, which typically was around 10 GPa. In highly frac- tured granodiorite, membrane failures were problematic. Commercially available pressuremeter devices for rock are currently limited to maximum pressures of around 30 MPa. Additional discussion of rock mass modulus is presented later in this chapter. TABLE 8 IN SITU TESTS WITH APPLICATIONS TO ROCK-SOCKET DESIGN Method Procedure Rock Properties Limitations/Remarks Pressuremeter (includes devices referred to as rock dilatometer) Pressuremeter is lowered to the test elevation in a prebored hole; flexible membrane of probe is expanded exerting a uniform pressure on the sidewalls of the borehole Rock mass modulus; rock mass strength in weak rocks ASTM D4719 Test affects a small area of rock mass; depending on joint spacing, may or may not represent mass behavior; limited to soft or weak rocks Borehole jack Jacks exert a unidirectional pressure to the walls of a borehole by means of two opposed curved steel platens Rock mass modulus; rock mass strength in weak rocks ASTM D4971 Measured modulus value must be corrected to account for stiffness of steel platens; test method can be used to provide an estimate of anisotropy Plate load test Load is applied to a steel plate or concrete foundation using a system of hydraulic jacks and a reaction frame anchored to the foundation rock Rock mass modulus; rock mass strength in weak rocks Loaded area is limited, so may not be effectively testing rock mass if joints are widely spaced; modulus values corrected for plate geometry, effect of rock breakage, rock anisotropy, and steel plate modulus; not common for deep foundations Texas cone penetration test Steel cone is driven by a drop hammer; number of blows per 300 mm of penetration is TCPT N-value; depth of penetration per 100 blows is penetration resistance (PR) Correlated to compressive strength of weak rocks encountered in Texas and Oklahoma Limitations similar to those of Standard Penetration Test; currently used by Texas and Oklahoma DOTs for direct correlation to side and base resistance of shafts in weak rock Notes: Adapted from Geotechnical Engineering Circular No. 5 (Sabatini et al. 2002). TCPT = Texas Cone Penetration Test.

23 An example of an in situ test that is used in a specific re- gion of the country is the Texas Cone Penetration Test (TCPT). A 76-mm-diameter solid steel cone is driven by a 77 kg (170 lb) drop hammer. The number of blows required to drive 300 mm (12 in.) is recorded and the results are given in one of two ways: (1) number of blows per 300 mm of pen- etration or TCPT N-value, or (2) the depth of penetration per 100 blows, referred to as the penetration resistance or PR. The Texas and Oklahoma DOTs use empirical correlations between the TCPT parameters and drilled shaft side and base resistances in soil and soft rock. The test procedure and correlations are available in the Texas DOT Geotechnical Manual, which can be accessed online. Some researchers have developed empirical correlations between TCPT mea- surements and properties of soft rock. For example, Cavu- soglu et al. (2004) show correlations between compressive strength of upper Cretaceous formation clay shales (UU tri- axial tests) and limestone (unconfined compression) and PR measurements conducted for Texas DOT projects. The cor- relations are highly formation-dependent and exhibit a high degree of scatter, but provide first order estimates of rock strength based on TCPT resistance in formations where sam- ple recovery is otherwise difficult. In addition to the tests identified as being applicable to rock, it is common practice to use in situ tests for soil to define the contact boundary between soil and rock. Of the agencies surveyed, 21 reported using the Standard Penetra- tion Test (SPT) and 3 reported using the Cone Penetration Test (CPT) to define the top-of-rock elevation. “Refusal” of the SPT or CPT penetration is the method most often used to identify rock. Limitations of this approach include the possi- bility of mistaking cobbles or boulders for the top-of-rock and the lack of consistency in SPT blowcounts in weak or weathered rock. Six states reported using the SPT in soft or weak rock to obtain rock properties (unconfined compressive strength) or for correlating SPT N-values directly to design parameters, principally unit side resistance. For example, the Colorado SPT-Based Method is used by the Colorado DOT to estab- lish design values of both unit side resistance and base resis- tance for shafts socketed into claystones when the material cannot be sampled in a way that provides intact core speci- mens adequate for laboratory uniaxial compression tests (Abu-Hejleh et al. 2003). O’Neill and Reese (1999) correlate unit side resistance with N-values for shafts in cohesionless IGMs, defined as materials with N > 50. Direct correlations between design parameters and N values are considered fur- ther in chapter three. Rock Mass Classification Several empirical classification systems have been proposed for the purpose of rating rock mass behavior. The most widely used systems are the Geomechanics Classification described by Bieniawski (1976, 1989) and the Rock Quality Tunneling Index described by Barton et al. (1974). Both sys- tems were developed primarily for application to tunneling in rock, but have been extended to other rock engineering problems. The application of classification systems to rock- socket design has been limited to correlations between clas- sification parameters and RMS and deformation properties. To facilitate such correlations, Hoek et al. (1995) introduced the GSI. Relationships were developed between GSI and the rock mass classifications of Bieniawski and Barton et al. The principal characteristics of the two classification systems are summarized, followed by a description of their relationship to GSI. For more detailed discussion, including limitations and recommended applications, consult the original refer- ences and Hoek et al. (1995, 2002). The Geomechanics Classification is based on determina- tion of the RMR, a numerical index determined by summing the individual numerical ratings for the following five cate- gories of rock mass parameters: • Strength of intact rock, • Drill core quality (in terms of RQD), • Spacing of discontinuities, • Condition of discontinuities, and • Groundwater conditions. An adjustment is made to the RMR for the degree to which joint orientation may be unfavorable for the problem under consideration. The classification system is presented in Table 9. Based on the RMR value, a rock mass is identified by one of five rock mass classes, ranging from very poor rock to very good rock. The draft 2006 Interim AASHTO LRFD Bridge Design Specifications recommends determination of RMR for classification of rock mass in foundation investiga- tions. Seventeen states reported using RMR either always or sometimes for rock mass classification associated with drilled shaft design. Barton and co-workers at the Norwegian Geotechnical In- stitute proposed a Tunneling Quality Index (Q) for describing rock mass characteristics and tunnel support requirements (Barton et al. 1974). The system is commonly referred to as the NGI-Q system or simply the Q-system. The numerical value of the index Q varies on a log scale from 0.001 to 1,000 and is defined as: (6) where RQD = rock quality designation, Jn = joint set number, Jr = joint roughness number, Ja = joint alteration number, Jw = joint water reduction factor, and SRF = stress reduction factor. Q J J J J n r a w = × × RQD SRF

Three states reported using the Q-system in connection with rock-socket design. A modified Tunneling Quality In- dex (Q') is utilized to determine the GSI, as described subsequently. The Geomechanics Classification can be used to estimate the value of GSI for cases where RMR is greater than 23, as follows: GSI = RMR89 – 5 (7) in which RMR89 is the RMR according to Bieniawski (1989) as presented in Table 9. For RMR89 values less than 23, the modified (Q′) is used to estimate the value of GSI, where: (8)Q J J Jn r a ' = × RQD 24 GSI = 9LogeQ' + 44 (9) Table 10 gives the values of the parameters used to evaluate Q' by Eq. 8. Engineering Properties of Rock Mass Shear Strength Geotechnical evaluation of foundation ultimate capacity un- der axial and lateral loading is calculated on the basis of shear strength along assumed failure surfaces in the rock or at the concrete–rock interface. Depending on the failure mode, the strength may need to be defined at one of three levels: (1) in- tact rock, (2) along a discontinuity, and (3) representative of a highly fractured rock mass. Figure 14 illustrates these cases for a socketed foundation in rock. For example, bearing A. Classification Parameters and Their Ratings (after Bieniawski 1989) Parameter Ranges of Values Strength of intact rock material Point load strength index, MPa >10 4–10 2–4 1–2 For this low range, uniaxial comp. test is preferred Uniaxial comp. strength, MPa >250 100–250 50–100 25–50 5–25 1–5 <1 1 Rating 15 12 7 4 2 1 0 Drill core quality, RQD (%) 90–100 75–90 50–75 25–50 <25 2 Rating 20 17 13 8 3 Spacing of discontinuities >2 m 0.6–2 m 200–600 mm 60–200 mm <60 mm 3 Rating 20 15 10 8 5 Condition of discontinuities Very rough surfaces, not continuous, no separation, unweathered wall rock Slightly rough surfaces, separation <1 mm, slightly weathered walls Slightly rough surfaces, separation <1 mm, highly weathered walls Slickensided surfaces or gouge <5 mm thick or joints open 1 to 5 mm continuous Soft gouge >5 mm thick or separation >5 mm continuous 4 Rating 30 25 20 10 0 Inflow per 10 m tunnel length None <10 10–25 25–125 >125 Ratio: Joint water pressure/ major principal stress 0 <0.1 0.1–0.2 0.2–0.5 >0.5 Ground- water General conditions Completely dry Damp Wet Dripping Flowing 5 Rating 15 10 7 4 0 B. Rating Adjustment for Joint Orientations Strike and dip orientations Very favorable Favorable Fair Unfavorable Very Unfavorable Ratings Foundations 0 –2 –7 –15 –25 C. Rock Mass Classes Determined from Total Ratings RMR 100 to 81 80 to 61 60 to 41 40 to 21 <20 Class Number I II III IV V Description Very good rock Good rock Fair rock Poor rock Very poor rock TABLE 9 GEOMECHANICS CLASSIFICATION SYSTEM FOR DETERMINATION OF ROCK MASS RATING (RMR)

25 capacity at the base of a socketed foundation in massive rock would be evaluated in terms of the strength of the intact rock. If the rock has regular discontinuities oriented as shown in level 2, base capacity may be controlled by the strength along the joint surfaces. If the rock is highly fractured (level 3), bearing capacity would have to account for the overall strength of the fractured mass. For each of the three cases, shear strength may be ex- pressed within the framework of the Mohr–Coulomb failure criterion, where shear strength (τ) is given by τ = c' + σ' tan φ' (10) in which c' = effective stress cohesion intercept, φ' = effec- tive stress angle of friction, and σ' = effective normal stress on the failure plane. Evaluation of shear strength for each of the three cases is summarized as follows. For intact rock the parameters c' and φ' can be determined from laboratory triaxial shear tests on specimens prepared from core samples. Triaxial testing procedures are given by ASTM D2664 and AASHTO T226. The survey of state DOTs indicates that triaxial testing is not used routinely. The most common test for intact rock is the uniaxial (unconfined) compression test, which can be considered a special case of triaxial testing with zero confining stress. The strength pa- rameter obtained is the uniaxial compressive strength, qu, which is related to the Mohr–Coulomb strength parameters by qu = 2c tan (45º + 1⁄2 φ) (11) However, the strength of intact rock is normally given simply in terms of qu. Stability analyses of rock sockets governed by massive rock are normally evaluated directly in terms of qu. When rock core is not sufficient for uniaxial compression testing, or sometimes for convenience, qu is correlated to re- sults of point load tests. Uniaxial compressive strength is also one of the parameters used for evaluating the strength of highly fractured rock masses, as discussed later. Shear strength of discontinuities can be determined using laboratory direct shear tests. The apparatus is set up so that the discontinuity surface lies in the plane of shearing between the two halves of the split box. Both peak and residual val- ues of the strength parameters (c' and φ') are determined. Discussion of direct shear testing of discontinuities, includ- ing its limitations, is given by Wyllie and Norrish (1996). For a planar, clean fracture (no infilling), the cohesion is zero and the shear strength is defined only by the friction angle. The roughness of the surface has a significant effect on the value of friction angle. If the discontinuity contains infilling, the strength parameters will be controlled by the thickness and properties of the infilling material. Compilations of typical representative ranges of strength parameter values for discon- tinuities are summarized in Mayne et al. (2001). The survey re- sults indicate that direct shear testing of joints is not conducted routinely by DOT agencies for rock-socket design. For intact rock masses and for fractured or jointed rock masses, Hoek and Brown (1980) proposed an empirical crite- rion for characterizing RMS. Since its appearance, this criterion has been applied widely in practice and considerable experience TABLE 10 JOINT PARAMETERS USED TO DETERMINE Q' 1. No. of Sets of Discontinuities = Jn 3. Discontinuity Condition & Infilling = Ja Massive 0.5 3.1 Unfilled cases One set 2 Healed 0.75 Two sets 4 Stained, no alteration 1 Three sets 9 Silty or sandy coating 3 Four or more sets 15 Clay coating 4 Crushed rock 20 3.2 Filled discontinuities Sand or crushed rock infill 4 2. Roughness of Discontinuities = Jr Stiff clay infilling <5 mm 6 Noncontinuous joints 4 Soft clay infill <5 mm thick 8 Rough, wavy 3 Swelling clay <5 mm 12 Smooth, wavy 2 Stiff clay infill >5 mm thick 10 Rough, planar 1.5 Soft clay infill >5 mm thick 15 Smooth, planar 1 Swelling clay >5 mm 20 Slick and planar 0.5 Filled discontinuities 1 *Note: Add +1 if mean joint spacing > 3 m. Modified from Barton et al. (1974). (a) Massive rock (b) Jointed rock (c) Highly fractured rock Shear failure along joint FIGURE 14 Base failure modes illustrating different operational shear strength conditions.

has been gained for a range of rock engineering problems. Based on these experiences, the criterion has undergone several stages of modification, most significantly by Hoek and Brown (1988), Hoek et al. (1995, 2002), and Marinos and Hoek et al. (2000). The nonlinear RMS is given by: (12) where σ'1 and σ'3 = major and minor principal effective stresses, respectively; qu = uniaxial compressive strength of intact rock; and mb, s, and a are empirically determined strength parame- ters for the rock mass. The value of the constant m for intact rock is denoted by mi and can be estimated from Table 11. Hoek and Brown σ σ σ 1 3 3 ' ' ' = + + ⎛ ⎝⎜ ⎞ ⎠⎟q m q su b u a 26 (1988) suggested that the constants mb, s, and a could be related empirically to the RMR described previously. Hoek et al. (1995) noted that this process worked well for rock masses with RMR greater than about 25, but not well for very poor rock masses. To overcome this limitation, the GSI was introduced. Sug- gested relationships between GSI and the parameters mb/mi, s, and a, according to Hoek et al. (2002) are as follows: (13) (14) (15) in which D is a factor that depends on the degree of disturbance to the rock mass caused by blast damage and stress relaxation. a e e= + − ⎛ ⎝⎜ ⎞ ⎠⎟ − −1 2 1 6 20 3 GSI 15 s D = − − ⎛⎝ ⎞⎠exp GSI 1009 3 m m D b i = − − ⎛⎝ ⎞⎠exp GSI 10028 14 Coarse Medium Fine Very fine Conglomerate (22) Sandstone 19 Siltstone 9 Claystone 4 Carbonate Breccia (20) Sparitic limestone (10) Micritic limestone 8 Chemical Gypstone 16 Anhydrite 13 Marble 9 Hornfels (19) Quartzite 24 Migmatite (30) Amphibolite 31 Mylonites (6) Gneiss 33 Schists (10) Phyllites (10) Slate 9 Granite 33 Rhyolite (16) Obsidian (19) Granodiorite (30) Dacite (17) Diorite (28) An desite 19 Gabbro 27 Dolerite (19) Basalt (17) Norite 22 Agglomerate (20) Breccia (18) Tuff (15) Light Dark Extrusive pyroclastic type Ig n eo u s Se di m en ta ry Non-foliated Slightly foliated Foliated*M et am or ph ic <------------ Graywacke --------------> (18) Clastic <--------------- Chalk -----------------> 7 <----------------- Coal -----------------> (8–21) Organic Non-clastic Rock Type Class Group Texture *These values are for intact rock specimens tested normal to foliation. The value of mi will be significantly different if failure occurs along a foliation plane. Note: Values in parentheses are estimates. TABLE 11 VALUES OF THE CONSTANT mi BY ROCK GROUP (Hoek et al. 1995)

27 The damage factor D ranges from zero for undisturbed in situ rock masses to 1.0 for very disturbed rock masses. Hoek et al. (2002) provide guidance on values of D for application to tun- nel and rock slope problems, but no work has been published relating D to drilled shaft construction. Some problems involving fractured rock masses (e.g., bear- ing capacity) are more readily analyzed in terms of the Mohr–Coulomb strength parameters than in terms of the Hoek–Brown criterion. Hoek and Brown (1997) noted that there is no direct correlation between the two sets of strength parameters. However, they describe a procedure that involves simulating a set of triaxial strength tests using the Hoek–Brown criterion (Eq. 12) then fitting the Mohr–Coulomb failure en- velope to the resulting Mohr’s circles by regression analysis. Values of the strength parameters c' and φ' defining the in- tercept and tangent slope of the envelope (which is nonlin- ear) can thus be determined. Hoek et al. (2002) presented the following equations for the angle of friction and cohesive strength of fractured rock masses: (16) (17) Applications of the Hoek–Brown criterion to rock-socket design are discussed further in chapter three (bearing capacity) and chapter four (lateral capacity). The draft 2006 Interim AASHTO LRFD Bridge Design Specifications rec- ommend the Hoek–Brown strength criterion for RMS char- acterization, but the earlier version (Hoek and Brown 1988) is presented rather than the updated version based on GSI. Deformation Properties Rock mass deformation properties are used in analytical methods for predicting the load-deformation behavior of rock-socketed foundations under axial and lateral loads. The parameters required by most methods include the modulus of deformation of the rock mass, EM, and Poisson’s ratio, v. Methods for establishing design values of EM include: • Estimates based on previous experience in similar rocks or back-calculated from load tests, • Correlations with seismic wave velocity propagation (e.g., Eqs. 1–3), • In situ testing, and • Empirical correlations that relate EM to strength or mod- ulus values of intact rock (qu or ER) and/or rock mass characteristics. Compilations of typical values of rock mass modulus and Poisson’s ratio are given in several sources, including c q a s a m s m a u b n b n a ' ' ' = +( ) + −( )[ ] +( ) +( −1 2 1 1 3 3 1 σ σ ) +( ) + +( ) +( ) +( ) − 2 1 6 1 2 3 1 a am s m a a b b n a σ ' φ' sin '= +( ) +( ) +( ) + + − − 1 3 16 2 1 2 6 am s m a a am s b b n a b σ mb n a σ '3 1( ) ⎡ ⎣⎢ ⎤ ⎦⎥− Kulhawy (1978), Wyllie (1999), and the AASHTO LRFD Bridge Design Specifications (2004). These values should be considered as general guidelines to expected ranges of values for different rock types and serve to illustrate the magnitude of variation that is possible. Rock mass modulus can vary from less than 1 MPa to greater than 100 GPa and depends on intact rock modulus, degree of weathering, and characteristics of discontinuities. Compiled values provide guidance for very preliminary evaluations, but should not be relied on for final design. Values of Poisson’s ratio exhibit a narrow range of values, typically between 0.15 and 0.3. Various authors have proposed empirical correlations between rock mass modulus and other rock mass proper- ties. Table 12 presents, in chronological order, some of the most widely cited expressions found in the literature. The earliest published correlations (expressions 1 and 2 of Table 12) relate EM to modulus of intact rock, ER, and RQD. In subsequent correlations (expression 3), RQD is replaced by RMR, providing a more comprehensive empirical ap- proach because six rock mass parameters (including RQD) are incorporated to evaluate the RMR. This was followed by correlations relating EM directly to rock mass indexes, including RMR and Q (expressions 4, 5, and 6). Hoek et al. (1995) show the graph given in Figure 15 with curves given by expressions 4, 5, and 6 of Table 12, along with case his- tory observations. The figure suggests that expression 4 of Table 12 provides a reasonable fit to the available data and offers the advantage of covering a wider range of RMR val- ues than the other equations. The draft 2006 Interim AASHTO LRFD Bridge Design Specifications recommend use of either expression 4 of Table 12 or a method recom- mended by O’Neill et al. (1996) based on applying a mod- ulus reduction ratio (EM/ER) given as a function of RQD in Table 13. Beginning with Hoek and Brown (1997), proposed corre- lation equations have been based on relating EM to GSI and properties of intact rock, either uniaxial compressive strength (qu) or intact modulus (ER). In expression 7, EM is reduced progressively as the value of qu falls below 100 MPa. This re- duction is based on the reasoning that deformation of better quality rock masses is controlled by discontinuities, whereas for poorer quality rock masses deformation of the intact rock pieces contributes to the overall deformation process (Hoek and Brown 1997). The version given in Table 12 is updated by Hoek et al. (2002) to incorporate the damage factor, D. The final correlation (expression 8) in Table 12 was pro- posed based on analyses by Yang (2006). Figure 16 shows a comparison of the regression equation (expression 8) to data from field observations of Bieniawski (1978) and Serafim and Pereira (1983), as well as modulus values measured by PMTs reported by Yang (2006). Expression 8 was applied to der- ivation of p-y curves for analysis of laterally loaded rock sockets, described further in chapter four. Additional discus- sion of empirical equations for rock mass modulus and their

application to foundation engineering is given by Littlechild et al. (2000), Gokceoglu et al. (2003), and Yang (2006). Rock mass modulus is a key parameter for rock-socket load-deformation analysis, which is a key step in the design process depicted in Figure 3. Several methods are identified in this chapter for establishing values of EM. These include geo- physical methods based on p-wave and s-wave velocities (Eqs. 1 and 2) or shear wave frequency (Eq. 3), in situ testing meth- ods (Table 8), and the correlation equations given in Table 12. The survey shows that correlation equations are the most widely used method for estimating modulus for rock-socket design, followed by in situ testing. The most common in situ test (used by five states) is pressuremeter (rock dilatometer), with a single state (Massachusetts) reporting use of the borehole 28 jack test. At least three other states using PMT for rock did not respond to the survey. The principal limitation of in situ testing is whether the volume of rock being tested is represen- tative of the in situ rock mass. Factors such as degree of rock disturbance, anisotropy, and spacing of discontinuities relative to the dimensions of the apparatus will determine the degree to which test results represent the response of rock mass to foundation loading. As noted earlier in this chapter, rock mass modulus measured by pressuremeter shows varying levels of agreement with other in situ testing methods. The full range of application and limitations of PMTs for rock mass modulus and its application to rock-socket design have yet to be deter- mined. Correlation equations for rock mass modulus have evolved over the years as illustrated by the relationships sum- marized in Table 12. Correlations are attractive because they are based on more easily measured properties of intact rock and rock mass indexes, but caution must be exercised because most of the correlations were developed specifically for appli- cations to tunneling. Calibration studies aimed at the applica- tion of correlation equations for rock mass modulus to load- deformation analysis of rock-socketed foundations are largely lacking at the present time. Studies by Littlechild et al. (2000) and Liang and Yang (2006) are exceptions and illustrate the type of additional work that is needed. TABLE 12 EMPIRICAL METHODS FOR ESTIMATING ROCK MASS MODULUS Expression Notes/Remarks Reference 1. EM = ER[0.0231(RQD) – 1.32] Reduction factor on intact rock modulus; EM/ER > 0.15 Coon and Merritt (1969); LRFD Bridge Design . . . (2004) 2. For RQD < 70: EM = ER (RQD/350) For RQD > 70: EM = ER [0.2 + (RQD – 70)/37.5] Reduction factor on intact rock modulus Bieniawski (1978) 3. += RMR4.111150 RMR1.0RM EE Reduction factor on intact rock modulus; EM/ER < 1.0 Kulhawy (1978) 4. 40 10 RMR 10)GPa( =ME 0 < RMR < 90 Serafim and Pereira (1983) 5. EM (GPa) = 2 RMR – 100 45 < RMR < 90 Bieniawski (1984) 6. EM (GPa) = 25 log10 Q 1 < Q < 400 Hoek et al. (1995) 7. 40 10 GSI 10 1002 1)GPa( = uM qDE for qu < 100 MPa 40 10 GSI 10 2 1)GPa( = DEM for qu > 100 MPa Adjustment to Serafim and Pereira to account for rocks with qu < 100 MPa; note qu in MPa Hoek and Brown (1997); Hoek et al. (2002) 8. 7.21GSI 100 e EE RM = Reduction factor on intact modulus, based on GSI Liang and Yang (2006) Notes: ER = intact rock modulus, EM = equivalent rock mass modulus, RQD = rock quality designation, RMR = rock mass rating, Q = NGI rating of rock mass, GSI = geological strength index, qu = uniaxial compressive strength. TABLE 13 ESTIMATION OF MODULUS RATIO (EM/ER) BASED ON RQD (O’Neill et al. 1996) EM/ER RQD (percent) Closed Joints Open Joints 100 1.00 0.60 70 0.70 0.10 50 0.15 0.10 20 0.05 0.05 RQD = rock quality designation. FIGURE 15 Rock mass modulus versus rock mass rating (Hoek et al. 1995).

29 A case history described by LaFronz et al. (2003) illustrates the use of multiple methods for establishing design values of rock mass modulus. Site characterization for the Colorado River Bridge (Hoover Dam Bypass Project) included borehole jack, downhole seismic (compression wave velocity), and lab- oratory uniaxial compression tests. The major rock unit for the abutment foundations on the Arizona side of the bridge is Hoover Dam tuff (welded volcanic ash). Results of field and laboratory tests used to establish rock mass modulus in the tuff are summarized in Table 14. Values given for the correlation with GSI reflect two values of GSI for the tuff, one corre- sponding to fracture conditions of width = 1 to 5 mm with soft filling (GSI = 45) and the other corresponding to fracture width of 0.1 to 1 mm and no filling (GSI = 52). Modulus values based on downhole p-wave velocities were calculated using equa- tions given by Viskne (1976), described by LaFronz et al. (2003) as valid at the rock mass scale. Results were applied as follows. Borehole jack measured values at stress ranges representative of expected footing bearing pressures were taken as reasonable values for de- veloping foundation load-deflection curves. Deformation modulus predicted by the correlation to GSI (Table 12, Hoek and Brown 1997) provided a cross-check on the borehole jack measured values. The mean value of modulus from the borehole jack tests is in the range of the GSI-predicted values. A low-strain modulus derived from downhole seis- mic measurements was used as a reasonable upper-bound check on the rock mass modulus. The modulus of intact rock from laboratory uniaxial compression tests on core samples is consistent with the observation of Heuze (1980) that field rock mass modulus values range from 20% to 60% of intact rock modulus and serve as an additional upper-bound check. INTERMEDIATE GEOMATERIALS A persistent challenge to the geotechnical engineer, and one that pertains directly to design and construction of drilled shafts, is defining the boundary between soil and rock. Dif- ferent approaches to site characterization and evaluation of geomaterial properties and different design methods are used when the geomaterial involved is clearly defined as soil or as rock. However, many geomaterials encountered in practice exhibit properties that make it difficult to define them clearly as being soil or rock within the context of standardized clas- sification systems. Geologic processes provide us with a con- tinuum of geomaterial properties and characteristics, some of which defy simplified categorization. The term intermediate geomaterial (IGM) has been ap- plied recently to earth materials with properties that are at the boundary between soil and rock (O’Neill et al. 1996). The criteria are based on (1) whether the material is cohe- sionless or cohesive and (2) some index of material strength. Cohesionless IGMs are defined by O’Neill et al. (1996) as very dense granular geomaterials, such as residual, completely decomposed rock and glacial till, with SPT N60-values between 50 and 100. Cohesive IGMs are defined as materials that exhibit unconfined compressive strengths in the range of 0.5 MPa ≤ qu ≤ 5 MPa. Specific materials identified by O’Neill et al. (1996) as being cohesive IGMs include (1) argillaceous geomaterials, such as heavily overconsoli- dated clays, clay shales, saprolites, and mudstones that are prone to smearing when drilled; and (2) calcareous rocks such as limestone and limerock and argillaceous geomaterials that are not prone to smearing when drilled. The term IGM as used by O’Neill et al. (1996) and subsequently adopted in O’Neill and Reese (1999) and in the draft 2006 Interim AASHTO LRFD Bridge Design Specifications has been limited specif- ically to design of drilled shafts and has not been adopted in the general geotechnical literature. For example, the term IGM is not used in the FHWA Manual on Subsurface Inves- tigations (Mayne et al. 2001) or in “Evaluation of Soil and Rock Properties,” Geotechnical Engineering Circular No. 5 (Sabatini et al. 2002). Responses to Question 8 of the survey show that most responding states (23) define IGMs for drilled shaft design according to the criteria of O’Neill et al. (1996). However, six states responded that geomaterials are classified as either soil or rock and IGM is not used. According to O’Neill and Reese (1999) cohesionless IGMs may be treated, for practical purposes, in the same Bieniawski (1978) Serafim and Pereira (1983) Ironton-Russell Regression FIGURE 16 Ratio of rock mass modulus to modulus of intact rock versus Geological Strength Index (Yang 2006). TABLE 14 MODULUS VALUES, HOOVER DAM TUFF (LaFronz et al. 2005) Method Mean Modulus (GPa) Borehole jack 2.83 Correlation to GSI 2.34, 3.52 Downhole seismic 3.31 Uniaxial compression 13.79

manner as coarse-grained (cohesionless) soils. They are as- sumed to respond to loading by rapid dissipation of excess pore water pressure (fully drained response) and are analyzed within the context of effective stress. For strength analysis, cohesionless IGMs are characterized in terms of the effective stress angle of friction φ'. It should be noted that some empirical correlations that apply to cohesionless soils, such as friction angle estimated from SPT N-values, may not be applicable to cohesionless IGMs. Specific approaches for estimating design parameters of shafts in cohesionless IGM are covered in chapter three. The definition of cohesive IGMs given earlier is based on a single index, the unconfined compressive strength. Al- though this categorization may be useful to identify mate- rials falling into a defined range of intact strength, it does not necessarily provide the distinction between soil and rock most relevant to behavior of drilled shafts. To illus- trate, consider Figure 17 from Kulhawy and Phoon (1993). This figure shows the relationship between unit side resis- tance determined from field load tests on drilled shafts and one-half of the unconfined compressive strength. Both pa- rameters are normalized by atmospheric pressure pa. Two categories of load tests were defined; those conducted on shafts in fine-grained soils (clay) and those in rock. Kul- hawy and Phoon relied on the judgment of the original authors and the database compilers to establish whether the material was soil or rock. For convenience, the range of normalized strength that defines cohesive IGM is superim- posed on Figure 17. It can be seen that the soil and rock data constitute apparently different populations, including over the range of strength that defines cohesive IGM. For pur- poses of drilled shaft side resistance, therefore, the classifi- cation of IGM does not provide a smooth transition from soil to rock. It may be more meaningful to define the mate- rial as being one or the other on the basis of additional geologic information. 30 There is no simple answer to the problem of classifying cohesive materials at the soil–rock boundary. Various classi- fications that distinguish geomaterials on the basis of com- pressive strength of unweathered rock material are summa- rized in Figure 18, which includes a proposed classification by Kulhawy et al. (1991) in which rock strength is defined relative to that of concrete used in construction, which is as- sumed to range from 20 kN/m2 (3 ksi) to 100 kN/m2 (15 ksi). Rock at the high end of the strength scale (>100 kN/m2) is classified as strong and in most cases would be expected to be an excellent founding material, except that it would be ex- pensive to excavate. Rock with compressive strength falling within the range of concrete strength is classified as medium and the rock mass could be either weaker or stronger than concrete, depending on weathering and structural features. For rock classified as weak (<20 kN/m2) foundation capacity is expected to be governed by the strength of the rock mass. Materials defined as cohesive IGMs by O’Neill et al. (1996) fall into this strength range. To account properly for the be- havior of weak rock in engineered construction, the follow- ing additional factors must be considered carefully: geologic origin, in situ weathering profile, state of stress, ground- water, and construction practices. A defining characteristic of geomaterials at the soil–rock boundary may be whether or not the in situ material was at one time rock (geologic origin). This is probably the distinguish- ing feature between clay and rock in Figure 17. The next geo- logic consideration is the in situ weathering profile. Igneous, sedimentary, or metamorphic rocks subjected to in-place weathering result in geologic profiles that may exhibit the full range of characteristics, for example, as described in Figure 11 (Key), Sheet 2, under “Rock Weathering—Alteration.” The descriptive terms are based on recommendations for describ- ing degree of weathering and alteration by the ISRM. One of the criteria for distinguishing between residual soil and com- pletely weathered or altered rock is whether the original rock FIGURE 18 Classification for unweathered rock material strength (Kulhawy et al. 1991). FIGURE 17 Side resistance versus geomaterial strength (Kulhawy and Phoon 1993). IGM

31 fabric is still apparent. The highest degree of weathering ap- plies to materials derived from rock but for which the rock fabric is not apparent. In this case, the material behavior is controlled by soil fabric and the material should be classified as residual soil, even though it may contain fragments of weathered rock. Materials in which the original minerals have been completely decomposed to secondary minerals but where the original fabric is intact may exhibit rock material behavior governed by rock mass features, including both rock material and discontinuities. The material should be consid- ered to be rock mass, even though it may be highly weathered or altered and exhibit low compressive strength. Judgment is always required in assessing whether material behavior is governed by soil fabric or by rock mass fabric; however, this is a key factor to be assessed in a design approach. Whether a geomaterial is assigned the term “IGM” or “weak rock” is not as important as understanding the geologic processes that give the material its characteristics and engineering properties. SUMMARY In this chapter, site characterization methods used to define subsurface conditions at bridge sites underlain by rock were reviewed. The survey shows that eight states currently use geo- physical methods to determine depth to bedrock and that seismic refraction is the method used. The literature review suggests that resistivity methods based on the use of multi- ple arrays can provide detailed profiles that may be useful for both design and construction. Karstic areas in limestone or dolomite terranes with irregular, pinnacled rock surfaces or solution cavities are examples of sites where recent de- velopments in geophysical methods could be applied. Every agency responding to the survey uses rock core drilling as the primary method of subsurface investigation for rock sockets. Current practice for description and classifica- tion of rock core is reviewed. The survey shows that most states routinely determine the RQD of rock core and that the uniaxial compressive strength of intact rock (qu) is also measured by one of the standardized methods. Also from the survey, it was determined that five states currently classify all rock mass according to the Geomechanics Classification System, in which rock mass is assigned a RMR. Twelve states use RMR occasionally, whereas 14 states indicated that RMR is never used. Some of the analytical methods de- veloped in recent years and described in subsequent chapters of this report require the rock mass classification in terms of RMR. Specifically, RMR can be used to evaluate strength parameters according to the Hoek–Brown failure criterion, a useful approach to quantifying strength of intact or highly fractured rock masses. RMR is used to establish GSI, which is required to use the most up-to-date version of the Hoek–Brown criterion. RMR and/or GSI are useful for esti- mating rock mass modulus using the empirical correlations given in Table 12. The RMR is also recommended in current FHWA manuals on site characterization and evaluation of soil and rock properties. Wider use of RMR or GSI classifi- cation of rock mass is one way that state DOT agencies can use the most up-to-date methods for characterizing RMS and deformation properties. In situ testing methods that provide information on rock mass modulus include PMT and borehole jack. Five states reported using these tests to obtain modulus values for rock- socket design. To use the best available analytical models for axial and lateral loading, as well as for effective interpretation of load test results, rock mass modulus is a required parameter. Currently, it is noted that there is no definitive in situ method or empirical equation for rock mass modulus that has been cal- ibrated specifically for application to design of rock sockets. A case history example is presented in this chapter illustrating the beneficial use of both in situ testing (borehole jack) and empirical correlations with GSI to establish representative val- ues of rock mass modulus for foundation design. Site and geomaterial characterization are interrelated with design, construction, and load testing of drilled shafts in rock. For design, Figure 3 shows that rock mass engineering properties required for analysis of rock-socket capacity and load-deformation response are obtained through field and laboratory testing. Table 15 is a summary of rock mass char- acteristics used in design methods for axial and lateral load- ing. A large X indicates the property is used directly in design equations that are currently applied widely in practice, whereas a small x indicates that the characteristic is used indirectly in the design or that it is required for a proposed design method that is not used widely. For example, intact rock modulus ER is not used directly to analyze load-displacement response of socketed shafts, but may be used to estimate the rock mass modulus EM, which is used directly in the analytical equations. Information obtained through the site investigation process will be used not only by design engineers but by contractors who will bid on the work and construct the foun- dations. As indicated in the flowchart shown in Figure 3, a goal of site characterization is to obtain information on constructability. O’Neill and Reese (1999) point out that con- tractors will be most interested in knowing the difficulties that might be encountered in drilling the rock. Specific in- formation that is useful in assessing the difficulty of drilling in rock includes loss or gain of drill water; rock type with lithological description; rock strength; characteristics of weathering; and rock mass characteristics such as the pres- ence, attitude, and thickness of bedding planes, foliation, joints, faults, stress cracks, cavities, shear planes, or other discontinuities. Boring logs, containing most of the informa- tion determined by the site investigation, are incorporated directly into the construction plans by most state DOTs. Any of the above information not given in the boring logs should be made available to bidders to facilitate informed decisions. The same information will be used by the design engineer to forecast potential construction methods and construction

problems to develop specifications for the project and to make cost estimates. Rock cores should be photographed and, when practical, retained for examination by prospective bidders. Field load testing, shown in the flowchart of Figure 3 and described in chapter five, provides direct verification of design assumptions regarding axial and lateral capacity and 32 load-deformation response. Results of field load testing also provide the basis for many of the design methods dis- cussed in the next two chapters. For correct interpretation of load test results, it is imperative that subsurface conditions and soil–rock engineering properties be evaluated as care- fully as possible. The properties required for design and listed in Table 15 are also required for load test interpretation and for proper extrapolation of load test results to design. TABLE 15 ROCK MASS ENGINEERING PROPERTIES REQUIRED FOR ROCK-SOCKET DESIGN Continuum Methods p-y Curve Parameter Compressive strength, intact rock, qu X X X x X Split tensile strength, intact rock, qt X Rock mass strength by Mohr– Coulomb or Hoek and Brown X x X Shear strength of joint surfaces X x Elastic modulus, intact rock, ER x x x Elastic modulus, rock mass, EM x X X X Rock quality designation (RQD) x X x x x x x x x Rock Mass Rating (RMR) X Geological Strength Index (GSI) X X Rock Mass Characteristic Axial Load- Displacement Unit Base Resistance Unit Side Resistance Design Applications Axial Loading Lateral Loading Load-Displacement Ultimate Resistance Notes: X = property is used directly in equations that are currently applied widely in practice. x = characteristic is used indirectly in the design or it is required for a proposed design method not widely used.

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TRB’s National Cooperative Highway Research Program (NCHRP) Synthesis 360: Rock-Socketed Shafts for Highway Structure Foundations explores current practices pertaining to each step of the design process, along with the limitations; identifies emerging and promising technologies; examines the principal challenges in advancing the state of the practice; and investigates future developments and potential improvements in the use and design of rock-socketed shafts.

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