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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
×
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Suggested Citation:"Chapter 3 - Field Load Testing." National Academies of Sciences, Engineering, and Medicine. 2011. Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. Washington, DC: The National Academies Press. doi: 10.17226/14574.
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14 Based on preliminary lateral pile load analyses, improvement of soft clay was considered to provide the greatest potential for increasing the lateral resistance of pile foundations. Therefore, a test site was selected where a variety of soil improvement methods appropriate for treating soft clay around a pile group could be investigated. The field load tests provided basic per- formance data that also could be used to calibrate and verify computer models. 3.1 Test Site Location The test site was located north of Salt Lake City, Utah, at the interchange of Redwood Road and I-215 on a Utah Depart- ment of Transportation right of way. An aerial view of the site is provided in Figure 3-1. The site offered several advantages including the following: 1. Presence of a consistent layer of relatively soft saturated cohesive soil near the ground surface, 2. Fill over the soft clay to allow easy access for construction equipment, 3. Access to water, and 4. Permission to drive piles and use ground improvement methods. Four pile groups were constructed at this site according to the basic layout shown in Figure 3-2 and discussed subsequently. 3.2 Geotechnical Site Characterization Geotechnical site conditions were evaluated using field and laboratory testing. Field testing included one drilled hole with undisturbed sampling, four cone penetration test (CPT) soundings, and shear wave velocity testing. Laboratory testing included unit weight and moisture content determination, Atterberg limits testing, and undrained shear testing. A generalized soil boring log at the test site is provided in Figure 3-3. The depth is referenced to the top of the excavation, which was 2.5 ft above the base of the pile cap as shown in the figure. The soil profile consists predominantly of cohesive soils; however, some thin sand layers are located throughout the profile. The cohesive soils typically classify as CL or CH materials with plasticity indices of about 20 as shown in Fig- ure 3-3(b). In contrast, the soil layer from a depth of 15 to 25 ft consists of interbedded silt (ML) and sand (SM) layers as will be highlighted by the subsequent plots of CPT cone tip resistance. The liquid limit, plastic limit, and natural moisture content are plotted in Figure 3-3(b) at each depth where Atterberg limit testing was performed. The water table is at a depth of 1.5 ft. The natural water content is less than the liquid limit near the ground surface, suggesting that the soil is overconsolidated, but the water content is greater than the liquid limit for soil speci- mens from a depth of 5 to 27 ft suggesting that these materials may be sensitive. Below a depth of 30 ft the water content is approximately equal to the liquid limit suggesting that the soils are close to normally consolidated. The undrained shear strength is plotted as a function of depth in Figure 3-3(c). Undrained shear strength was mea- sured using a miniature vane shear test or Torvane test on undisturbed samples immediately after they were obtained in the field. In addition, unconfined compression tests were per- formed on most of the undisturbed samples. Both the Torvane and unconfined compression tests indicate that the undrained shear strength decreases rapidly from the ground surface to a depth of about 6 ft but then tends to increase with depth. This profile is typical of a soil profile with a surface crust that has been overconsolidated by desiccation. However, the undrained shear strength from the unconfined compression tests is typi- cally about 30% lower than that from the Torvane tests. The unconfined compression tests at a depth of 27 and 48 ft appear to have been conducted on soil with sand lenses because the measured strength is substantially lower than that from the C H A P T E R 3 Field Load Testing

Torvane test and are not likely to be representative of the soil in-situ. The undrained shear strength was also computed from the cone tip resistance using the following correlation equation: where qc is the cone tip resistance, σ is the total vertical stress, and Nk is a variable which was taken to be 15 for this study. The undrained shear strength obtained from Equation 2 also is plotted versus depth in Figure 3-3(c) and the agreement with the strengths obtained from the Torvane and unconfined com- pression tests is reasonably good. Nevertheless, there is much greater variability and the drained strength in the interbedded sand layers is ignored. A summary of laboratory test results is provided Table 3-1. Four cone penetration tests were performed across the test site and plots of cone tip resistance, friction ratio, and pore pressure are provided as a function of depth in Figure 3-4. In addition, the interpreted soil profile also is shown. From the ground surface to a depth of about 15 ft the soil profile appears to be relatively consistent with a cone tip resistance of about 6 tons per square foot (tsf) and a friction ratio of about 1%. However, one thin sand layer is clearly evident between 6 and 8 ft. The cone tip resistance, friction ratio, and pore pressure plots clearly show the interbedded silt and sand layering in the soil profile between 15 and 27 ft. below the ground surface. s q N u c k = −( )σ ( )2 Figure 3-5 provides plots of the cone tip resistance, friction ratio, and pore pressure versus depth as a function of depth for all four of the CPT soundings. The measured parameters and layering are generally very consistent for all four soundings, which indicates that the lateral pile load tests can be fairly com- pared from one site to the next. Figure 3-6 provides a plot of the shear wave velocity as a function of depth obtained from the downhole seismic cone testing. The interpreted soil profile and cone tip resistance are also provided in Figure 3-6 for reference. The shear wave veloc- ity in the upper 10 ft of the profile is between 300 and 400 ft/sec, which is relatively low, and suggests a low shear strength. Between a depth of 10 to 20 ft, the velocity increases to about 550 ft/sec. This increase in velocity is likely associated with the interbedded layer that contains significant sand layers. Below 20 ft, the velocity drops to a value of around 500 ft/sec and remains relatively constant to a depth of 45 ft. 3.3 Single Pile Test in Untreated Soil Test Pile Properties A 12.75-in. OD pipe pile with a 0.375-in. wall thickness was driven closed-ended with a hydraulic hammer to a depth of 45 ft below the excavated ground surface on June 15, 2007. The test pile had a beveled end that allowed a 1.5-in. thick 15 N Test Site (150 ft x 40 feet approx.) Silt Fence Figure 3-1. Aerial view of the test site and surrounding area.

Figure 3-2. Generalized layout of test pile groups at the test site. Virgin Mass Mixing5ft x 9ft x 10ft Jet Grouting 5ft x 9ft x 10ft Flowable Fill 9ft x 9ft x 5ftVirgin Soil Flowable Fill 5ft x 9ft x 5ft Jet Grouting 9ft x 9ft x 10ft VirginVirgin Compacted Fill Compacted Fill5ft x 9ft x 5ft Geopiers 5ft x 9ft x 10ft Pile Cap 1Pile Cap 2Pile Cap 3Pile Cap 4 32 ft9 ft 9 ft 9 ft 9 ft32 ft 32 ft N

17 Figure 3-3. Borehole log, plot of Atterberg limits and natural water content vs depth, and plot of undrained shear strength vs depth. Table 3-1. Summary of laboratory soil test data. 0 5 10 15 20 25 30 35 40 45 50 0 20 40 60 (b) Moisture Content (%) D ep th B el ow E xc av at io n (ft ) PL LL wn 0 5 10 15 20 25 30 35 40 45 50 0 20 40 60 (a) Soil Profile D ep th B el ow E xc av at io n (ft ) LEAN CLAY (CL) w/ Sand Lenses FAT CLAY (CH) w/ Sand Lenses LEAN CLAY (CL) w/ Silt Lenses SANDY SILT (ML) with Silty Sand and Lean Clay Layers SANDY LEAN CLAY (CL) w/ Sand Lenses LEAN CLAY (CL) w/ Sand Lenses Interbedded LEAN CLAY (CL) and SANDY SILT (ML) 0 5 10 15 20 25 30 35 40 45 50 0 250 500 750 1000 1250 (c) Undrained Shear Strength, su (psf) D ep th B el ow E xc av at io n (ft ) Unconfined Torvane CPT In-Place Atterberg Limits Depth (ft) Dry Unit Weight d (lb/ft3) Natural Moisture Content wn (%) Liquid Limit (LL) (%) Plastic Limit (PL) (%) Plasticity Index (PI) (%) Unconfined Compressive Strength (lb/ft2) Miniature Vane Shear Strength (Torvane) (lb/ft2) Unified Soil Classification System Symbol 1.25 117.6 34.2 39 18 21 1104 - CL 2.75 117.4 34.4 38 18 20 626 620 CL 5.75 104.6 56.0 51 21 30 384 320 CH 8.5 112.4 41.5 38 18 20 684 534 CL 11.5 110.8 44.1 38 19 19 741 500 CL 16.5 126.6 24.2 19 18 1 1081 560 ML 26.75 116.9 35.0 27 14 13 237 780 CL 33.5 124.6 26.1 27 14 13 1306 780 CL 36.75 117.1 34.8 35 17 18 1381 840 CL 41.75 112.0 42.1 46 17 29 1037 520 CL 48 117.2 34.6 33 16 17 297 660 CL plate to be welded flush with the edge of the pile at the bottom. The steel conformed to ASTM A252 Grade 3 specifications and had a yield strength of 58,700 psi based on the 0.2% offset criteria. The moment of inertia of the pile itself was 279 in.4; however, angle irons were welded on opposite sides of the test pile, which increased the moment of inertia to 342 in.4. A steel reinforcing cage was installed at the top of the test pile to replicate the reinforcing cages in the test piles within the pile groups. The reinforcing cage consisted of six #8 reinforcing bars that were confined within a #4 bar spiral with a diameter of 8 in. The reinforcing cage extended to a depth of 10 ft. The steel pipe pile was filled with concrete that had an average unconfined compressive strength of 5150 psi based on tests of four specimens. A drawing of the cross-section for the test pile is provided in Figure 3-7. Test Layout, Instrumentation, and Procedure The lateral load test was conducted on October 10, 2007, after the pile had been in the ground for about 4 months. The ground around the test pile was excavated to the elevation of

05 10 15 20 25 30 35 40 45 50 0 42 6 Friction Ratio, Rf (%) D e p t h B e l o w E x c a v a t i o n ( f t ) CPT 2 0 5 10 15 20 25 30 35 40 45 50 0 100 200 300 Pore Pressure, u (ft) D e p t h B e l o w E x c a v a t i o n ( f t ) CPT 2 Hydrostatic 0 5 10 15 20 25 30 35 40 45 50 0 20 40 60 Soil Profile D e p t h B e l o w E x c a v a t i o n ( f t ) LEAN CLAY (CL) w/ Sand Lenses FAT CLAY (CH) w/ Sand Lenses LEAN CLAY (CL) w/ Silt Lenses SANDY SILT (ML) with Silty Sand and Lean Clay Lenses SANDY LEAN CLAY (CL) w/ Sand Lenses LEAN CLAY (CL) w/ Sand Lenses Interbedded LEAN CLAY (CL) and SANDY SILT (ML) 0 5 10 15 20 25 30 35 40 45 50 0 50 100 150 Cone Tip Resistance, qt (tsf) D e p t h B e l o w E x c a v a t i o n ( f t ) CPT 2 Figure 3-4. Plots of cone tip resistance, friction ratio and pore pressure vs depth curves from cone penetration test (CPT) Sounding 2 near the center of the site along with soil profile.

05 10 15 20 25 30 35 40 45 50 0 2 4 6 Friction Ratio, Rf (%) D e p t h B e l o w E x c a v a t i o n ( f t ) CPT 1 CPT 2 CPT 3 CPT 4 0 5 10 15 20 25 30 35 40 45 50 0 100 200 300 Pore Pressure, u (ft) D e p t h B e l o w E x c a v a t i o n ( f t ) CPT 1 CPT 2 CPT 3 CPT 4 Hydrostatic 0 5 10 15 20 25 30 35 40 45 50 0 20 40 60 Soil Profile D e p t h B e l o w E x c a v a t i o n ( f t ) LEAN CLAY (CL) w/ Sand Lenses FAT CLAY (CH) w/ Sand Lenses LEAN CLAY (CL) w/ Silt Lenses SANDY SILT (ML) with Silty Sand and Lean Clay Lenses SANDY LEAN CLAY (CL) w/ Sand Lenses LEAN CLAY (CL) w/ Sand Lenses Interbedded LEAN CLAY (CL) and SANDY SILT (ML) 0 5 10 15 20 25 30 35 40 45 50 0 50 100 150 Cone Tip Resistance, qt (tsf) D e p t h B e l o w E x c a v a t i o n ( f t ) CPT 1 CPT 2 CPT 3 CPT 4 Figure 3-5. Plots of cone tip resistance, friction ratio, and pore pressure vs depth curves from all CPT soundings at the site along with soil profile.

20 0 5 10 15 20 25 30 35 40 45 50 0 20 40 60 Soil Profile D ep th B el ow E xc av at io n (ft ) LEAN CLAY (CL) w/ Sand Lenses FAT CLAY (CH) w/ Sand Lenses LEAN CLAY (CL) w/ Silt Lenses SANDY SILT (ML) with Silty Sand (SM) and Lean Clay (CL) Layers SANDY LEAN CLAY (CL) w/ Sand Lenses LEAN CLAY (CL) w/ Sand Lenses Interbedded LEAN CLAY (CL) and SANDY SILT (ML) 0 5 10 15 20 25 30 35 40 45 50 0 200 400 600 800 1000 Shear Wave Velocity (fps) D ep th B el ow E xc av at io n (ft ) 0 5 10 15 20 25 30 35 40 45 50 0 50 100 150 Cone Tip Resistance, qT (tsf) D ep th B el ow E xc av at io n (ft ) Figure 3-6. Plots of cone tip resistance and shear wave velocity versus depth from seismic cone testing along with soil profile. Direction of Loading 12.75 inch OD pipe pile with 0.375 in wall thickness (fy=58.6 ksi) 6-#8 longitudinal bars (fy=60 ksi) with 8 inch diameter #4 bar spiral at 4 inch pitch Concrete in-fill (f'c=5150 psi) 1.5"x1.5"x0.25" angles (fy=36 ksi) (rotated 18 from line of loading) Figure 3-7. Cross-section of single pipe pile.

the base of the pile caps used in the pile group testing, which was approximately 2.5 ft below the ground surface shown in Figure 3-9. Load was applied at a height of 1.5 ft above the sur- rounding ground surface. In contrast to the pile group tests where the pile head was restrained (fixed-head), the pile head for the single pile test was unrestrained (free-head). A free- head lateral load test for a single pile is common because it is very difficult to create a truly fixed-head condition for a single pile. Because the purpose of the test was to calibrate the analy- sis model, it was considered more important to know the pile head fixity condition than whether it was a free-head or fixed- head. The boundary condition can be changed easily for sub- sequent analyses once the soil model is established. The load was applied to the test pile using a hydraulic jack attached to an electric pump. Hemispherical plates were used to prevent the application of moment to the pile and account for any eccentricity in the loading. Load was measured using a resistance-type strain gauge load cell that had been calibrated previously in the laboratory. Pile head deflection was measured at the elevation of the load point with a string potentiometer attached to an independent reference frame. In addition, pile head deflection was measured at an elevation 3.31 ft above the load point so that pile head rotation could be computed. Prior to placing concrete in the test pile, a 1-in. diameter conduit was installed to a depth of 30 ft. A shape accelerometer array was inserted into this conduit at the beginning of the load test so that deflection versus depth profiles could be determined at various load increments. Data was recorded using computer data acquisition systems. A photograph of the test pile during testing is provided in Figure 3-8. The load test was performed incrementally using a deflec- tion control approach. The load in the hydraulic jack was increased to deflection increments of 0.125, 0.25, 0.50, 0.75, 1.0, 1.5, 2.0, and 2.5 in. The maximum deflection was some- what larger than that used for the pile group testing to facili- tate calibration of the numerical models. After reaching each target deflection, the deflection was maintained for 3 minutes and then load was reduced to zero prior to loading the pile to the next increment. 21 Figure 3-8. Photograph of lateral load test on single pipe pile. Figure 3-9. Complete pile head load vs pile head deflection curve. 0 5 10 15 20 25 30 35 0.0 0.5 1.0 1.5 2.0 2.5 3.0 Pile Head Deflection (in) Pi le H e a d Lo ad (k ip s )

Test Results A plot of the complete pile head load versus deflection curve for the entire test is presented in Figure 3-9. This curve provides the load path taken during loading, unloading, and reloading for each cycle. While the load was decreased to zero after each increment, the pile did not return to its initial zero deflection level, but exhibited a residual deflection. This may have been due to side friction and soil falling into the gap behind the pile. During reloading, the load-defection curve was stiffer than that observed during virgin loading at the same deflection. The virgin pile head load versus deflection curve is plotted in Figure 3-10. This plot was developed by plotting the peak values and eliminating the unload and reload segments. The curve exhibits the conventional hyperbolic shape that would be expected for a pile in soft clay. The peak pile head load versus rotation curve is also plotted in Figure 3-11. The rotation, θ, was determined using the following equation: where Δ1 is the pile deflection 3 ft above the load point, Δ2 is the pile deflection at the load point, and H is the distance between the measurements (3.31 ft). Deflection versus depth curves obtained from the shape accelerometer arrays are provided in Figure 3-12 for a number θ = −( )⎛⎝⎜ ⎞ ⎠⎟−tan ( )1 1 2 3 Δ Δ H 22 Figure 3-10. Peak pile head load vs pile head deflection. 0 5 10 15 20 25 30 35 0.0 0.5 1.0 1.5 2.0 2.5 3.0 Pile Head Deflection (in) Pi le H ea d Lo ad (ki ps ) Pipe Pile 0 5 10 15 20 25 30 35 0.0 0.2 0.4 0.6 0.8 1.0 1.2 1.4 Pile Head Rotation (Degrees) Pi le H ea d Lo ad (k ips ) Pipe Pile Figure 3-11. Peak pile head load vs pile head rotation.

of deflection increments. The shape array provides horizon- tal deflection values at 1 ft intervals from the top of the pile, which was approximately 40 in. above the load point. Without any corrections, the computed deflection curves obtained from the shape arrays are consistent with the max- imum pile head deflections measured by the string poten- tiometers at the load point. The deflected shape curves also are consistent with the free-head (zero-moment) boundary condition. 3.4 Pile Group Properties A total of 16 lateral load tests were performed on the 4 pile groups. Schematic drawings of the pile group layout and the soil improvement geometries are provided in Appendix A. All pile groups consisted of nine test piles, which were driven in a 3 × 3 arrangement with a nominal center to center spacing of 3 ft. The tests piles were 12.75-in. outside diameter pipe piles with a 0.375-in. wall thickness and they were driven closed-ended with a hydraulic hammer to a depth of approx- imately 44 ft below the excavated ground surface. The steel conformed to ASTM A252 Grade 3 specifications and had a yield strength of 58.6 ksi based on the 0.2% offset criteria. The moment of inertia of the pile itself was 279 in.4; however, angle irons were welded on opposite sides of two to three test piles within each group, which increased the moment of inertia to 324 in.4. A steel reinforcing cage was installed at the top of each test pile to connect the test piles to the pile cap. The reinforcing cage consisted of six #8 reinforcing bars that were confined within a #4 bar spiral with a diameter of 8 in. and a pitch of 6 in. The test piles typically extended about 2 ft above the base of the pile cap and the reinforcing cage extended 2.25 ft above the base of the cap and 8.75 ft below the base. The steel pipe pile was filled with concrete that had an average uncon- fined compressive strength of 5000 psi. A pile cap was constructed by excavating 2.5 ft below the surface of the surface clay layer. The concrete was poured directly against vertical soil faces on the front and back sides of each pile cap. This procedure made it possible to evaluate pas- sive force against the front and back faces of the pile caps. In contrast, plywood forms were used along the sides of each cap and were braced laterally against the adjacent soil faces. This construction procedure created a gap between the cap side- wall and the soil so that side friction would be eliminated. Steel reinforcing mats were placed in the top and bottom of each cap. 3.5 Pile Group Testing Procedure Lateral load was applied using MTS actuators with the load centered at a height of 11 in. above the top of the pile cap. Each actuator could produce 600 kips in compression and 450 kips in tension. Another pile group or groups provided a reaction for the applied load. In all cases, the reaction pile group or groups were located 32 ft away from the test pile group to minimize interference between the two pile groups during lateral load- ing. Each actuator was fitted with two 8.67-ft extension pieces to span the 32.1-ft gap between the pile groups. The actuator was attached to a concrete corbel atop each pile cap using steel tie-rods that extended through PVC sleeves in the corbel and were bolted to the back face of the corbel. This allowed load to be applied without affecting the soil around the pile group. The tie-rods were prestressed to minimize displacement of the steel during the load tests. A three-dimensional swivel head was located at each end of the actuator to provide a zero moment or “pinned” connection. Each swivel could accommodate ± 5° of pile cap rotation about a horizontal line and ± 15° of pile cap rotation about a vertical line. The lateral load tests were carried out with a displacement control approach with target pile cap displacement increments of 0.25, 0.5, 0.75, 1.0, and 1.5 in. During this process, the actu- ator extended or contracted at a rate of about 40 mm/min. Additionally, at each increment, 10 cycles with a peak pile cap amplitude of ±0.1 in. were applied with a frequency of approx- imately 1 Hz to evaluate dynamic response of the pile cap. After this cyclic loading at each increment, the pile group was pulled back to the initial starting point prior to loading to the next higher displacement increment. 23 Figure 3-12. Deflection vs depth curves at several deflection increments for single pile lateral load test. -5 0 5 10 15 20 -1.0 0.0 1.0 2.0 3.0 4.0 D e pt h fro m G ro u n d Su rf a c e (ft ) Horizontal deflection (in) 0.5 in 0.75 in 1.0 in 1.5 in 2.0 in 2.5 in String Pot Above Load Pt. String Pot At Load Pt.

Load Test Instrumentation Applied load was measured directly by the load cell on the actuator, which was calibrated in the laboratory prior to test- ing in the field. Lateral pile cap displacement was measured using two string potentiometers attached to the pile cap at the elevation of the loading point (0.92 ft above the top of the cap) on the east and west sides of the actuator attachment point as shown in Figure 3-13. Lateral pile cap displacement also was measured on the back side of each corbel with two string potentiometers attached 1.75 ft (21 in.) and 0.375 ft (4.5 in.) above the top of the pile cap directly in line with the load direc- tion. Therefore, the vertical distance between these two string pots was 1.375 ft (16.5 in.) as shown in Figure 3-13. Finally, ver- tical pile cap displacement was measured at two points along the length of each pile cap to evaluate pile cap rotation. On both caps, string potentiometers were located 2 in. from the north and south edges of the corbel, with a distance of 44.72 in. between the potentiometers on Cap 1 and a distance of 108 in. (9 ft) for Cap 2 as shown in Figure 3-13. Each potentiometer was attached to an independent reference beam supported at a distance of about 6 ft from the side of the pile cap. The pile rotation, θ, was determined using the following equation: where Δ1 and Δ2 are the vertical pile cap deflection at two points on the pile cap and H is the distance between the measurements. θ = −⎛⎝⎜ ⎞⎠⎟−tan ( )1 1 2 4 Δ Δ H 24 Figure 3-13. Typical instrumentation layout for piles caps with a partial-length corbel (Caps 1 and 4) and a full-length corbel (Caps 2 and 3).

Prior to placing concrete in the test piles, a 1-in. diameter PVC pipe was installed to a depth of 30 ft in the middle pile in each row of each pile group. A shape accelerometer array could be inserted into this pipe at the beginning of the load tests so that deflection versus depth profiles could be determined at various load increments. Using triaxial accelerometers embed- ded into a flexible cable at 1-ft intervals, the shape arrays pro- vided real-time displacement versus depth profiles throughout the process of testing. To provide some check on the accuracy of the shape array measurements, inclinometer pipes were also installed in the middle pile in the front and back rows of each pile group. Inclinometer measurements were typically per- formed before testing and then again once the 1.5-in. or final displacement increment had been reached. Bending moment along the length of the piles was evaluated using two comple- mentary procedures. First, the deflection versus depth curves obtained from the shape array data were used to determine bending moment versus depth profiles along the length of the pile. The moment, M, was computed using the following equation: where E is the elastic modulus of the pile; I is the moment of inertia of the pile; y −1, yo, and y1 are the horizontal pile deflections at locations 12 in. above, at the depth, and 12 in. below the depth of interest; and h is the vertical spacing between the deflection (12 in.). For the steel pipe pile with concrete fill, this required a cal- culation of the composite properties. These calculations indicated that EI was 1.415 × 107 kip-in2 using a compressive strength of 5100 psi based on compression tests on concrete cylinders at the time of testing. The moment computed using Equation 5 is very sensitive to minor variations or errors in the measured displacement versus depth curves. To reduce the influence of minor variances in the measured displacement data on the computed moment, a 5th-order polynomial equa- tion was developed based on the measured data to smooth the displacement versus depth curves. The displacements used in Equation 5 were then based on values computed with the polynomial equation. Although the difference in the displace- ment values at any depth were generally very small, this pro- cedure produced moment versus depth curves with more realistic shapes. Secondly, waterproof electrical resistance type strain gauges were placed at depths of 2, 6, 11, and 13.5 ft below the top of two to three piles within each group. For Pile Cap 1, the mid- dle piles within each row were instrumented with strain gauges while for Pile Cap 2, the middle piles within the front and back M EI y y y h = − +( ) −1 0 1 2 2 5( ) row piles were instrumented. The strain gauge depths were selected to provide the maximum negative and positive moments along the pile. For a fixed-head or restrained-head pile, the maximum negative moment is expected to occur at the pile-pile cap interface. Preliminary LPILE analyses suggested that the maximum positive moment would likely occur between 11 and 13 ft below the top of the piles. Angle irons were welded on opposite sides of the instrumented piles to a depth of 20 ft to protect the strain gauges during pile driving. Data was recorded using two computer data acquisi- tion systems. 3.6 Pile Group Tests in Untreated Clay Plan and profile drawings showing the layout of the pile group in untreated clay for Tests 1 and 2 are provided in Fig- ure 3-14. Tests 1 and 2 were performed to provide a baseline of the lateral load behavior of the pile caps in virgin soil con- ditions prior to any soil treatment. Test 1 was conducted by pulling the caps together using the actuator while the untreated native soil was in place adjacent to the pile cap. At the comple- tion of Test 1, the pile cap was pulled back to zero deflection, but after the actuator load was released some residual deflec- tion remained. Prior to Test 2, the soil immediately adjacent to the opposite face of the pile cap was excavated by hand to cre- ate roughly a 1-ft-wide gap between the pile cap face and the adjacent soil as shown in Figure 3-14. This excavation elimi- nated passive force against the pile cap for the subsequent test. After excavation was complete, which required less than an hour to accomplish, Test 2 was carried out by pushing the pile caps apart using the actuator. The testing was performed using the same procedure described previously. Test 2 was designed to define the passive force provided by the unsaturated clay soil against the pile cap. Load versus Displacement Plots of the complete pile cap load versus displacement curves for Cap 1 are provided in Figure 3-15. This plot provides the load path taken during loading, unloading, and reloading for each cycle. At the end of each loading cycle it was necessary to apply a tensile force to bring the actuator deflection back to zero. This does not appear to be a result of yielding in the pile based on measured moments. The behavior could result from a flow of weak soil into the gap behind the pile during loading or lateral resistance due to side shear on the pile as it moves in the opposite direction. During reloading, the load is typically less than that obtained during virgin loading and considerably more linear, but after the load exceeds the maximum previ- ous load, the load increase and the load deflection transitions into what appears to be the virgin curve. 25

NFigure 3-14. Plan and profile drawings of Pile Caps 1 and 2 during Test 1 when the pile groups were pulled together by the actuator. (During Test 2, the soil adjacent to the pile cap was excavated to the base of the cap and the pile caps were pushed apart by the actuator.)

The virgin pile head load versus displacement curves for each pile group have been developed in Figure 3-16 by plotting the peak values and eliminating the unload and reload seg- ments. The curve exhibits the conventional hyperbolic shape that would be expected for a pile in soft clay. Despite the fact that the two pile groups are 32 ft apart and have minor varia- tions in construction details, the two load-displacement curves are nearly identical. These results suggest that the soil prop- erties across the site are sufficiently uniform for valid com- parisons to be made between the pile caps with various soil improvement techniques relative to the untreated conditions. Rotation versus Load Pile cap rotation versus load curves based on the string potentiometer and shape arrays for Cap 1 are provided in 27 -200 -150 -100 -50 0 50 100 150 200 250 300 350 -0.5 0 0.5 1 1.5 2 Displacement (in) Lo ad (k ip s) 0 50 100 150 200 250 300 350 0 0.5 1 1.5 2 Displacement (in) Lo a d (ki ps ) T1 Cap 1 T1 Cap 2 Figure 3-15. Complete pile cap load vs pile head deflection curve for Cap 1 during Test 1. Figure 3-16. Peak pile cap load vs pile head deflection curves for Caps 1 and 2 during Test 1.

28 Figure 3-17. Peak pile cap rotation vs load for Caps 1 and 2 during Test 1. Figure 3-18. Deflection vs depth curves at several deflection increments for Pile Cap 1 during Test 1. Array 106 Test 1 Cap 1 North Pile 0 5 10 15 20 25 30 Horizontal displacement (in) D e pt h Fr o m T op o f C o rb e l (f t) Array 104 Test 1 Cap 1 Middle Pile 0 5 10 15 20 25 30 -0.5 0.0 0.5 1.0 1.5 2.0 -0.5 0.0 0.5 1.0 1.5 2.0 Horizontal Displacement (in) D e pt h Fr o m T op o f C o rb e l (f t) 0.125 in 0.5 in 0.75 in 1.0 in 1.5 in Average String Pot 0.125 in 0.5 in 0.75 in 1.0 in 1.5 in Average String Pot 1-S A-142 Load A-104 A-106 1-M 1-N 1-S A-142 Load A-104 A-106 1-M 1-N Figure 3-17. The curves are fairly linear up to a load of about 170 kips after which the rotation begins to increase more rapidly with load. The measured rotations are fairly consis- tent for both caps, although the rotation of Cap 1 is some- what greater than that for Cap 2. Although pile cap rotation is clearly observed, it is considerably lower than the rotation of the single pile under free-head conditions. Displacement versus Depth Curves Displacement versus depth curves obtained from the shape accelerometer arrays in the piles within Pile Cap 1 are provided in Figure 3-18. One of the shape arrays in each pile cap appeared to be providing unrealistic information, which was likely due to damage from previous field testing. As a result, profiles are only provided for two shape arrays in each cap. The location of the shape arrays relative to the piles in the group and the loading direction are shown by the legends in each fig- ure. The average displacements measured by the string poten- tiometers at the elevation of the load application for each load increment are also shown in these figures for comparison pur- poses. The displacements obtained from the shape arrays are generally quite consistent with those measured by the string potentiometers; however, in some cases, variations are observed. The discrepancies appear to be related to the difficulty of providing a tight fit between the shape array and the surrounding PVC pipe in some cases. The deflected shape curves are generally consistent with a restrained-head bound- ary condition. Some rotation is observed, but the rotation is small relative to a free-head pile subjected to the same load lev- els (see single pile test results). It appears that the shape arrays were long enough to extend below the depth where lateral dis- placements dropped off to zero.

Figure 3-18 provides comparisons between the displace- ment versus depth curves obtained from the shape arrays and the two inclinometer pipes in Pile Cap 1. Since inclinometer soundings were only taken at the maximum displacement, comparisons are only provided for one increment. Because the inclinometer soundings required 20 minutes to perform the displacement profiles from the shape arrays are sometimes dif- ferent than the values for the 1.5-in. displacement increments shown in Figure 3-19. The displacement profiles from the shape arrays are quite consistent with the profiles from the inclinometers. These results provide increased confidence in the accuracy of the profiles. It should be noted, however, that the inclinometer profiles, which extend deeper into the pile, indicate that some negative displacement is occurring below the base of the shape arrays. Maximum Moment versus Load Curves Figures 3-20 and 3-21 provide plots of the maximum nega- tive and positive bending moments versus applied pile cap load, respectively, for Cap 2 during Test 1. Moment data come from both shape array and strain gauge data when available. Initially, the curves are relatively linear; however, the bending moment tends to increase more rapidly with load at the higher load levels as soil resistance is overcome. The curves from the strain gauges provide relatively consistent moment versus load curves with little evidence of strong group interaction effects for the displacement levels involved. The agreement between the curves computed by the strain gauges and shape arrays varies. Test 2 Results after Excavation Adjacent to Pile Cap As previously indicated, the two pile caps were pulled back to zero displacement at the end of Test 1. However, when the load was released, the caps relaxed back toward the direction they had previously been pushed leaving a residual (negative) displacement offset of about 0.3 in. at the start of Test 2. Because the pile caps during Test 2 were pushed in the oppo- site direction to those from Test 1, the residual deflection is given a negative sign. Figure 3-22 provides a comparison between the load-displacement curves for Caps 1 and 2 during Tests 1 and 2. The load-displacement curves for Test 2 have been shifted right slightly (0.15 in.) to account for gap effects so that the curve for Cap 2 matches the curves for Caps 1 and 2 during Test 1 at larger displacements than would be expected. A comparison of load-displacement curves for Cap 1 with and without passive force on the pile cap can then be made and the results indicate that the passive force is approximately 50 kips. Based on the curves in Figure 3-22, the passive force versus displacement curve shown in Figure 3-23 has been developed, 29 Figure 3-19. Comparison of displacement vs depth curves measured by shape arrays and inclinometers for Cap 1 during Test 1. Array 106 Test 1 Cap 1 Inclinometer Comparison 0 5 10 15 20 25 30 35 40 45 Horizontal Displacement (in) D e pt h Fr om T op o f C o rb e l (f t) Final North Array Final North Inclinometer Array 104 Test 1 Cap1 Inclinometer Comparison 0 5 10 15 20 25 30 35 40 45 -0.5 0.0 0.5 1.0 1.5 2.0 -0.5 0.0 0.5 1.0 1.5 2.0 Horizontal Displacement (in) D e pt h Fr om T op o f C o rb e l (f t) Final Middle Array Final North Inclinometer Final South Inclinometer 1-S A-142 Load A-104 A-106 1-M 1-N 1-S A-142 Load A-104 A-106 1-M 1-N

30 (a) Test 1 Maximum Negative Moments in Pile 1-N 0 10 20 30 40 50 60 70 80 0 50 100 150 200 250 300 Maximum Load (kips) M a x im u m M o m en t (k ip - ft) Strain Gage 1-S Load A-106 1-M 1-N A-104A-142 (b) Test 1 Maximum Negative Moments in Pile 1-M 0 50 100 150 200 250 300 0 20 40 60 80 100 120 Maximum Load (kips) M ax im u m M o m en t ( ki p- ft) Strain Gage Array 104 Load 1-S 1-M 1-N A-106A-104A-142 (c) Test 1 Maximum Negative Moments in Pile 1-S 0 50 100 150 200 250 300 Maximum Load (kips) 0 10 20 30 40 50 60 70 80 M ax im u m M o m en t ( ki p- ft) Strain Gage Load 1-S 1-M 1-N A-106A-104A-142 Figure 3-20. Maximum negative moment vs total pile cap load for Piles (a) 1-N, (b) 1-M, and (c) 1-S in Cap 2 during Test 1.

31 (a) Test 1 Maximum Positive Moments in Pile 2-N (b) Test 1 Maximum Positive Moments in Pile 2-M (c) Test 1 Maximum Positive Moments in Pile 2-S 0 10 20 30 40 50 60 70 80 0 50 100 150 200 250 300 Maximum Load (kips) M ax im u m M om en t ( ki p- ft) Strain Gage Array 134 Load 2-S 2-M 2-N A-112 A-115 A-134 0 50 100 150 200 250 300 Maximum Load (kips) 0 10 20 30 40 50 60 70 80 M ax im u m M om en t ( ki p- ft) Array 115 Load 2-S 2-M 2-N A-112 A-115 A-134 0 50 100 150 200 250 300 Maximum Load (kips) 0 20 40 60 80 100 120 M ax im u m M o m en t ( kip -ft ) Strain Gage Load 2-S 2-M 2-N A-112 A-115 A-134 Figure 3-21. Maximum positive moment vs total pile cap load for piles (a) 2-N, (b) 2-M, and (c) 2-S in Cap 2 during Test 1.

32 0 50 100 150 200 250 300 350 -0.5 0 0.5 1 1.5 2 Displacement (in) Lo ad (k ip s ) T2 Cap 1 T2 Cap 2 T1 Cap 1 T1 Cap 2 0 50 100 150 200 0 0.5 1 1.5 2 Displacement (in) Pa s s iv e Fo rc e (k ips ) Figure 3-22. Comparison of peak pile cap load vs pile head deflection curves for Caps 1 and 2 during Tests 1 and 2. Figure 3-23. Interpreted passive force vs deflection curves based on comparison of Tests 1 and 2 on Pile Cap 1. which indicates that the full passive force was essentially devel- oped with a displacement of about 0.75 in. Additional test results for Test 2 are provided in Adsero (2008, Appendix 2). 3.7 Pile Group Load Tests Involving Jet Grouting Plan and profile views of the jet grout columns around Pile Caps 1 and 2 are shown in Figure 3-24. Jet grouting treatment for Pile Cap 1 involved treatment adjacent to the pile group. Treatment for Pile Cap 2 involved treatment below and around the pile group. A single-hole double fluid jet grouting tech- nique was employed to form the grout columns and each of the columns was constructed with identical installation param- eters. The jet grout drill head was initially advanced to the base of the treatment zone using water jets and a drill bit located at the bottom of the drill rod. Subsequently, the drill head was rotated and pulled upward at a constant rate, while cement slurry was injected at a specified pressure and flow rate from the inner orifice of the drill nozzle. Concurrently, compressed air was injected from the outer orifice of the drill nozzle to form a protective shroud around the slurry jet to improve the

NFigure 3-24. Plan and profile views of Pile Groups 1 and 2 after treatment with jet grouting.

erosive capacity of the cement slurry jet. The grout slurry mix had a specific gravity of 1.52, which is equivalent to a 1:1 water to cement ratio by weight. Jet Grout Treatment below Pile Cap 2 A total of eight 5-ft diameter soilcrete columns were installed beneath and around Pile Cap 2 to a depth of 10 ft below the bot- tom of the pile cap. Four of the columns were installed at the periphery of the pile cap, and an additional four were installed through the cap itself. During construction of the pile cap, four 6-in. diameter PVC pipes were placed in the pile cap between the rebar to allow easy access to the jet grout pile after construc- tion. After constructing the cap, backfill soil was placed over the cap to allow the jet grout rig to move over the cap. Four PVC pipes were extended to the ground surface to provide the jet grout drill rod with an unobstructed path through the fill material and the pile cap. For retrofit projects these access holes would have to be drilled through the pile cap. The jet grout columns were spaced at approximately 3 ft center-to-center in the north-south direction and 5 ft center-to-center in the east- west direction. This likely produced a 2-ft overlap between columns in the north-south direction, but there was little or no overlap between columns in the east-west direction. The grout treatment extended about 3 ft beyond the front and back ends of the cap and somewhat beyond the cap on the top and bottom sides. Each of the columns was constructed with identical instal- lation parameters. These parameters are summarized in Table 3-2. One rotation of the high-pressure nozzles occurred in a 0.11-in. lift. Based on the column diameter, flow rates, pull rates, and rotation rates, the cement content for the jet grout columns would be expected to be about 26 lbs/ft3 or about 20% by weight. Jet Grout Treatment Below Cap 1 A total of seven soilcrete columns were installed in two rows to create a wall along one edge of the foundation. Plan and pro- file views of the jet grout columns adjacent to Pile Cap 1 are shown in Figure 3-24. The target diameter of each of the columns was 4 ft and they were spaced 3-ft-on-center in a tri- angular pattern. This created an overlap between columns of approximately 1 ft. Each jet grout column extended from the top of the pile cap to a depth of 12 ft below the top of the pile cap. The centers of the first row of jet grout columns were positioned so that the jet could cut underneath the pile cap and produce a soilcrete wall that would intersect the front row of piles. Based on the target column diameter, the soil- crete columns likely extended about 1.5 ft under the pile cap, or about to the middle of the outside row of piles. Each of the columns was constructed using identical con- struction parameters that are summarized in Table 3-3. One rotation of the high-pressure nozzles occurred in a 0.14-in. lift. Based on the column diameter, flow rates, pull rates, and rota- tion rates, the cement content for each jet grout column would be expected to be about 24 lbs/ft3 or about 20% cement by weight. Compressive Strength Testing of Jet Grout Columns Wet grab samples were taken from five completed columns below Cap 2 and two columns adjacent to Cap 1. The samples were taken from locations near the top, middle, and bottom of the columns. In addition, core samples were taken from the top of two columns adjacent to Cap 1 a few weeks after treatment. Prior to testing, the cored samples measured 4 in. in diameter with an approximate length to diameter ratio of 2.0. The unconfined compressive strength of the soilcrete pro- duced by the jet grouting process was evaluated using the wet grab samples as well as core samples. Figure 3-25 provides a summary of the compressive strength test results as a function of time after treatment. Although there is significant scatter to the data, which is typical for soilcrete columns installed using jet grouting, there is a trend of increasing strength with curing time. Although the compressive strength of the untreated soil prior to treatment was approximately 4 psi, the average com- pressive strength after jet grout treatment reached about 680 psi with mean ±1 standard deviation bounds ranging from about 34 Column Length 10 ft Estimated Column Diameter 5 ft Grout Pressure 6000 lbs/in.2 Grout Flow Rate 90 gallons/min Rotation Speed 7 revolutions/min Pull Rate 0.79 in./min Column Length 12 ft Estimated Column Diameter 4 ft Grout Pressure 6000 psi Grout Flow Rate 90 gallons/min Rotation Speed 8 revolutions/min Pull Rate 1 in./min Table 3-2. Jet grouting installation parameters for columns created beneath Pile Cap 2. Table 3-3. Jet grouting installation parameters for columns installed adjacent to Pile Cap 1.

500 to 800 psi. These strength gains are typical for jet grouting applications in similar soils (Burke, 2004). Ground improve- ment specialty contractors typically use a design value of about one-third the value measured from field test specimens to account for variations in properties within the treated zone. Using this approach, the compressive strength of the jet grouted zone would be about 250 psi; however, even cored specimens had strengths of 480 psi as shown in Figure 3-25. The average strength from two cored samples is about 30% lower than the strength obtained from the wet grab samples. The strength from the core samples is likely more representative of in-situ conditions and is attributable to the poorer mixing produced by the jet grouting process relative to the hand mixing employed with the wet grab samples. Test Results for Cap 1 (Jet Grouting Adjacent to Cap) Treating the soil adjacent to Pile Cap 1 with jet grouting increased the lateral resistance of the pile cap substantially. The results from Test 3 and Test 6 were combined to create a com- posite load-displacement curve for the pile cap following jet grouting. The combined curve is presented in Figure 3-26. The 35 Figure 3-25. Compressive strength of jet grout columns as a function of time after treatment along with design strength values typically employed by geotechnical specialty contractors. 0 200 400 600 800 1000 1200 0 5 10 15 20 25 30 35 40 Curing Time (Days) St re n gt h (p si ) Column 1 Column 2 Column 3 Column 6 Column 7 Column 10 Column 14 Cored Samples Mean + Std Dev. Mean - Std Dev. Mean Wet-Grab Sample Strengths Design Strength Range Core Sample combined curve had a maximum load of 612 kips at a pile cap displacement of 0.72 in., which is 398 kips greater than the 214 kip maximum load from the virgin curve for the same dis- placement. This represents an increase in lateral resistance of 185% at the maximum measured deflection. Tests were also performed on Pile Cap 1 after the soil adja- cent to the pile cap had been excavated. Despite excavation of the soil, the load-displacement curve was essentially the same after consideration of reloading effects (typically a 10% reduc- tion). Although the soilcrete mass was not connected to the pile cap, it was connected to the piles below because jet grouting extended under the cap. Therefore, lateral movement of the piles engaged the soilcrete mass and produced the same lateral resistance. Test Results for Pile Cap 2 (Jet Grouting below the Cap) Figure 3-27 presents a plot of the load-displacement curve for Pile Cap 2 after jet grouting in comparison with the virgin load-displacement curves. Comparing the resistance at a dis- placement of 1.5 in., jet grouting increased the lateral pile cap resistance from 282 kips to nearly 782 kips. This increase of

500 kips equates to an increase in total resistance of about 2.6 times or 160%. In addition, the initial stiffness of the load- displacement curve after jet grouting is considerably higher than the initial stiffness during virgin loading. The pile cap only displaced 0.016 in. at a load of 200 kips. The load-displacement curve can be separated into three distinct parts. The initial 0.3 in. of the curve are fairly linear. At a displacement of about 0.3 in. the curve shows an abrupt change in slope. A sec- ond, relatively linear portion of the curve extends from 0.3 to about 1.6 in. of displacement. The third portion of the curve following 1.6 in. of displacement is flat with a slight drop off in strength after 2.1 in. of displacement. This shape is much dif- ferent than the hyperbolic shape of the load-displacement curve for the virgin tests and is likely associated with the differ- ent deflections required to mobilized adhesive resistance on the soilcrete mass (0.25 in.) relative to that for passive force (2 in.). The load-displacement curve after excavation to a depth of 7 ft in front of the jet grout zone is also shown in Figure 3-27. Initially, the stiffness is not much greater than that for the pile group in untreated virgin soil; however, the ultimate resis- tance exceeds 600 kips at a displacement of 3 in. To produce a more readable report, additional plots, simi- lar to those presented for the pile group in virgin clay are not presented here but are available in Adsero (2008, Appendix 2). 3.8 Pile Group Load Tests Involving Soil Mixing Construction Details Plan and profile drawings of the pile group with a soil mix wall on one side of Pile Cap 1 are provided in Figure 3-24. The soil mixed wall adjacent to the cap was 10 ft deep, 11 ft wide, and extended 4 ft in front of the cap. Because of the small size of the wall, economics did not permit the mobilization of a dedicated soil mixing rig to the site. Instead, a procedure was applied to produce a volume of soil with a compressive strength and consistency typical of that produced by soil mix- ing. The native soil was first excavated to a depth of 5 ft below the top of the cap using a trackhoe. The excavation was then filled to the top of the cap with jet grout spoils from the oppo- site side of the cap. Afterward, the remaining intact soil from 36 0 100 200 300 400 500 600 700 800 900 -0.5 0 0.5 1 1.5 2 2.5 3 3.5 4 Displacement (in) Lo ad (ki ps ) Virgin Virgin - Excavated Jet Grout Jet Grout - Excavated Figure 3-27. Combined load-displacement curves for tests performed on Pile Cap 2 following jet grouting. The results from the virgin test also are shown for comparison. 0 50 100 150 200 250 300 350 400 450 500 550 600 650 -1 -0.5 0 0.5 1 1.5 2 Displacement (in) Lo ad (k ips ) T1 Cap 1 Virgin T2 Cap 1 Virgin - Excavated Cap 1 Jet Grout Figure 3-26. Combined load-displacement curves for all tests performed on Pile Cap 1 following jet grouting. The results from the virgin test also are shown for comparison.

5 to 10 ft below the top of the cap was progressively excavated with the excavator bucket and mixed with the jet grout spoils. Mixing was accomplished by repeatedly stirring the native soil and grout spoil until the consistency of the mixture became relatively homogeneous and no large blocks were obvious in the mixture. This process required approximately 10 to 15 minutes of mixing and provided a 1:1 ratio of soil to grout spoil mixture. The grout used in the jet grouting procedure was designed to have a specific gravity of approximately 1.52, which is the equivalent of a 1:1 water to cement ratio by weight using nor- mal Type I cement. The cement content per volume of jet grout slurry was computed to be about 24 lbs/ft3. Mixing the jet grout slurry with the underlying clay at a 1:1 ratio by vol- ume reduced the cement content of the resulting soilcrete wall to approximately 12 lbs/ft3. This corresponds to about 10% cement by weight. Six core samples obtained from the soil- crete wall indicate that the mean compressive strengths were 130 and 140 psi after 30 and 60 days of curing, respectively. This strength gain is consistent with past experience for soil mixed walls (Terashi, 2003). Test Results for Pile Cap 1 with Soil Mixing Figure 3-28 presents plots of the load-displacement curves for Cap 1 in untreated virgin clay and Test 3 after the mass mix soil improvement. With the soil mix wall, the pile cap resisted 452 kips compared to the 282 kips resisted by the pile cap in the virgin clay at a displacement of 1.5 in. This represents an increase of 60% in the lateral resistance provided by the pile cap. It also is interesting to evaluate the increase in initial stiff- ness due to the mass mixing. Prior to treatment, the secant stiffness of the load-displacement curve at a displacement of 0.1 in. was 800 kips/in.; after soil mixing the stiffness increased to 1300 kips/inch. This represents an increase in stiffness of about 62%. Figure 3-29 provides plots of the load-displacement curves for Cap 1 during Test 2 in virgin clay and Test 4 after the mass mix wall construction. In contrast to Tests 1 and 3, in these two tests the soil adjacent to the cap was excavated to the base of the cap. Because the soil had been previously loaded, the load- displacement curve for the pile cap with the mass mix wall is actually lower than that for the pile cap in virgin clay. However, as displacement increases to the maximum previous displace- ment, the load-displacement curve appears to follow the load- displacement curve for the pile cap in virgin clay with little apparent increase. Because the soil adjacent to the pile cap had been excavated, the pile cap no longer pushed the soil-mixed wall laterally; hence, no increase in lateral resistance was pro- duced. This behavior is in contrast with that for Pile Cap 2 where jet grouting allowed the soilcrete to extend underneath the pile cap and contact the piles themselves. To produce a more readable report, additional plots, simi- lar to those presented for the pile group in virgin clay are not presented here but are available in Herbst (2008, Appendix 3). 3.9 Pile Group Load Tests Involving Flowable Fill Several sets of lateral load tests were performed after exca- vating and replacing the soil around Pile Cap 3 with flowable fill. One set of tests was performed for the case where the soil below the pile cap was excavated and replaced with flowable fill prior to driving the test piles. The technique would repre- sent an approach for improving lateral resistance for new construction. Plan and profile drawings for this case are shown in Figure 3-30. In this case the flowable fill extended 37 Figure 3-28. Comparison of measured load- displacement curves for Cap 1 in virgin soil and after construction of a soil mix wall on one side of the cap. Figure 3-29. Comparison of measured load- displacement curves for Cap 1 in virgin soil and after construction of a soil mix wall on one side of the cap followed by excavation of soil adjacent to the cap. 0 100 200 300 400 500 0 0.5 1 1.5 2 Displacement (in) Lo ad (k ip s) Soil Mixing-T9 Cap 1 Virgin Soil-T1 Cap1 0 100 200 300 400 -0.5 0 0.5 1 1.5 2 2.5 3 3.5 Displacement (in) Lo ad (k ip s) Virgin Excavated T2 Cap 1 Soil Mix Excavated T15 Cap 1

Figure 3-30. Plan and profile views of Cap 3 (right) and Cap 4 (left) during Tests 3 and 5.

directly below the cap to a depth of 6 ft below the base of the pile cap. The fill was flush with the edge of the pile cap on one side but extended 5 ft beyond the pile cap on the other side to evaluate the effect of improvement width in front of the pile groups. The flowable fill zone was originally intended to be deeper, but the depth had to be reduced to prevent failure of the excavation in the weak clay layers. Although the flowable fill was designed to have an unconfined compressive strength of about 100 psi, only one of the six test cylinders was intact enough to be tested and that test cylinder only had a compres- sive strength of 30 psi. Therefore, the flowable fill was proba- bly closer to a weakly cemented sand at a medium relative density. For Test 3, the two pile groups were pushed apart but for Test 5 the pile groups were pulled together. Because of the lower than expected compressive strength of the original flowable fill zone, a second set of lateral load tests was subsequently performed after constructing a flowable fill wall adjacent to the pile cap. This technique would represent an approach for improving lateral pile group resistance after construction. Plan and profile drawings for this case are pro- vided in Figure 3-31. The flowable fill zone was only 6 ft deep, 12 ft wide, and extended 6 ft in front of the pile cap. The flow- able fill was designed to have a compressive strength of 150 psi; however, the average of four test cylinders was 137 psi. For Test 10, the pile groups were pushed apart but for Test 12 the pile groups were pulled together. Some concern has been expressed about long-term strength loss of flowable fill in saturated conditions with groundwater flow. Therefore, three flowable fill cylinders were kept in a fog room and tested 700 days after placement. The test results for these cylinders were very consistent and yielded an average compressive strength of 57 psi, which represents a 56% decrease in strength over 2 years’ time. Visual observations of the test cylinders did indicate that indeed some of the cemen- titious material had leached out. Leaching was observed as white streaks on the outside of the samples. The leaching occurred because water was able to flow into the flowable fill. If the flowable fill had higher cement content the leaching would have been reduced. Also, if the water did not flow over the flowable fill, the leaching would have been reduced or possibly eliminated. Load Test Results Figure 3-32 shows the load-displacement curves for Cap 3 during Test 3 after treatment with flowable fill compared to Test 1, Cap 2 in untreated virgin soil with soil to the top of the pile cap. In these tests, the pile caps were both in contact with the adjacent soil. Both curves have the same general hyperbolic shape; however, the flowable fill treatment increased the resis- tance by about 20 to 30 kips or about 10% relative to the pile cap in untreated soil. Figure 3-33 provides a plot of the load- displacement curves for Cap 3 during Test 5 after treatment with flowable fill compared to Cap 1 during Test 2 in untreated virgin soil. In both cases, the soil adjacent to the pile cap was excavated. The ultimate resistance for the cap with flowable fill was once again about 30 kips greater than that for the pile cap in untreated clay. This represents an increase of about 10% rel- ative to the pile cap in untreated clay. These results indicate that excavating the weak clay and replacing it with the weakly cemented sand provided only minimal increases in lateral resistance. Figure 3-34 provides a comparison of the lateral load- displacement curves for Cap 3 for Test 10 where the flowable fill extended to the top of the pile cap and for Test 12 where a 1-ft wide slot was excavated to the base of the pile cap imme- diately adjacent to the cap. The flowable fill wall increased the lateral resistance at a displacement of 1.85 in. by about 150 kips. This represents an increase in lateral resistance of about 50% with relatively little cost or effort. Figure 3-35 provides a comparison of the lateral load-dis- placement curve for Cap 3 after excavation of the slot rela- tive to the curve for Cap 1 in untreated clay after excavation adjacent to the cap. The load-displacement curves for both cases are relatively comparable, suggesting that the increase in resistance was achieved when the pile cap impacted the flowable fill wall and caused it to move into the surround- ing ground. As the wall moved laterally, both passive force on the back of the wall and adhesive resistance on the side of the wall could produce increased lateral resistance. When the slot was excavated next to the cap, the cap did not impact the wall and the resistance was about the same as that for the cap in untreated clay. The load test results for the flowable fill wall are very simi- lar to those obtained for the soil mixed wall and suggest that the mechanism of increased resistance is produced by passive force and adhesive shear on the side walls as the wall is pushed into the surrounding soil rather than by increased lateral pile- soil resistance. The results also suggest that the treated zone may only need to have an unconfined compressive strength of 140 psi to effectively behave as a “rigid wall” in developing increased lateral resistance. To produce a more readable report, additional plots, similar to those presented for the pile group in virgin clay are not presented here but are available in Miner (2009, Appendix 3). 3.10 Pile Group Load Tests Involving Excavation and Replacement Excavation and Replacement with Compacted Fill Plan and profile drawings showing the layout of Pile Cap 4 with compacted fill are provided in Figure 3-36. Tests on this 39

Figure 3-31. Plan and profile views of Cap 3 (left) and Cap 2 (right) during Tests 10 and 12. Test 10 performed with flowable fill adjacent to pile cap and Test 12 performed after excavation of flowable fill adjacent to cap.

41 Figure 3-32. Load vs displacement results comparing Test 3 on Cap 3 (weak flowable fill below the cap) to Test 1 on Cap 2 (untreated clay). Figure 3-33. Load vs displacement curves for Test 5 on Cap 3 (weak flowable fill below the cap excavated to base of cap) to Test 2 on Cap 1 (untreated clay excavated to base of cap). 0 50 100 150 200 250 300 350 0 0.5 1 1.5 2 Displacement (in) Lo ad (k ip s) Weaker flowable fill beneath the cap with passive T3 Cap 3 untreated clay with passive T1 Cap 2 0 50 100 150 200 250 300 350 -0.25 0.25 0.75 1.25 1.75 2.25 Displacement (in) Lo ad (k ip s) untreated clay w / o passive T2 Cap 1 Weaker f lowable fill beneath the cap w / o passive T5 Cap 3

pile group were designed to determine the increased strength that could be provided by excavating the soft clay and replac- ing it with compacted sand. Prior to pile driving, clay was exca- vated to a depth of 6.25 ft and replaced with compacted fill up to the base of the pile cap. Clean concrete sand, meeting ASTM C-33 specifications, was used as the backfill material. The sand was compacted in 6- to 8-in. lifts using a hydraulic plate com- pactor attached to the end of a trackhoe. Based on nuclear den- sity measurements, the sand was compacted to an average in-place dry density of 104.2 lb/ft3, which is 93.7% of the mod- ified Proctor density (γd max = 111 lbs/ft3). Plans originally called for excavation and replacement to greater depth; however, cav- ing of the soft clay precluded deeper excavation. When the piles were installed, the ground heaved and, in order to maintain the correct pile cap thickness, approximately 0.75 ft of backfill had to be removed, leaving approximately 3 ft of sand under the cap. The sand fill extended 5 ft beyond the cap face on one side to evaluate the increased pile-soil resistance from extending the 42 0 50 100 150 200 250 300 350 400 450 0 0.5 1 1.5 2 Displacement (in) Lo ad (k ips ) Flowable fill behind cap w/o passive T12 Cap 3 Flowable fill behind cap with passive T10 Cap 3 Figure 3-34. Load vs displacement results comparing Tests 12 and 10. Figure 3-35. Load vs displacement results comparing Test 2 on Cap 1 to Test 12 on Cap 3.

Figure 3-36. Plan and profile views of Pile Caps 3 and 4 after excavation and replacement with compacted fill around Pile Cap 4 and placement of flowable fill under Pile Cap 3.

sand fill. Lateral load tests were performed in both directions. Comparison with the pile caps in Tests 1 and 2 allow a deter- mination of the increased resistance for sand fill. Test Results for Compacted Fill A comparison of the load-displacement curves for Tests 1 and 5 is provided in Figure 3-37. Test 1 involves the pile cap in untreated clay; compacted sand was placed directly below the pile cap for Test 5. The comparison shows an increase in lateral resistance of about 23 kips at a displacement of 1.5 in. resulting from placing compacted fill directly below the pile cap. This represents an 8% increase in resistance relative to the total resistance from soil-pile interaction and passive force or a 10% increase in resistance relative to soil-pile resis- tance alone. Figure 3-38 provides a comparison of the load-displacement curves for Test 3 relative to Test 2. In contrast with Test 5 where the compacted fill stopped at the end of the pile cap, the com- pacted fill extends 5 ft beyond the end of the cap for Test 3. For both Tests 2 and 3 the soil adjacent to the pile cap was exca- vated so no passive resistance was present in either test. A com- parison of the two curves indicates that the compacted fill increased the lateral soil-pile resistance by about 40 kips. As expected, extending the compacted fill 5 ft beyond the cap increased the lateral resistance; however, the increase was rela- tively small. The increased resistance represents an increase of 18% relative to a comparable pile group in untreated clay. This increase in lateral resistance can only be attributed to increased soil-pile resistance because there was no soil adjacent to the pile cap. The increase of 18% is comparable to results reported by Brown et al. (1986, 1987) when a stiff clay was replaced with compacted sand at a relative density of 50%. Greater improvement could potentially have been achieved if the compacted fill could have extended deeper; however, this would have required flatter excavation slopes to prevent cav- ing and more backfill material, which would increase the cost. Finite element studies conducted by Weaver and Chitoori (2007) suggest that most of the benefit from compacted fill around a pile occurs for fill materials extending five pile dia- meters below the ground surface based on FEM analysis. In this case, the fill extended about three pile diameters. Figure 3-39 provides a comparison of the load-displacement curves for Test 3 and Test 4. The only difference between the two tests is that for Test 4 sand was compacted adjacent to the pile cap extending 5 ft beyond the cap. Therefore, the differ- ence between the two tests represents the passive force that the 5-ft wide and 2.5-ft thick layer of sand produced. A compari- son between the load-displacement curves for Tests 3 and 4 at the greatest displacements indicates that the ultimate passive force with the sand backfill was approximately 32 kips. This passive force is actually less than the 50-kip passive force mea- sured when the native clay was left in place adjacent to the pile cap face in Test 1, as discussed previously. This decrease in pas- sive force occurs because the native clay in the upper 2.5 ft of the profile is desiccated and relatively strong. However, if the clay in the upper 2.5 ft of the profile were softer, excavation and replacement with compacted sand could have increased the passive force. For example, if the clay surface layer had an undrained shear strength of only 500 psf, the passive force in the clay would only have been about 25 kips. Rammed Aggregate Pier Construction Rammed aggregate piers (RAPs) are a shallow alternative to deep foundations. They create a dense gravel column that 44 Figure 3-37. Load displacement comparison of Test 1 with Test 5 (shifted to the right 0.4 in.). 0 50 100 150 200 250 300 350 -0.2 0 0.2 0.4 0.6 0.8 1 1.2 1.4 1.6 Displacement (in) Lo ad (k ip s) Untreated Clay T1 Cap 1 Compacted Sand Under Cap T5 Cap 4

reinforces the surrounding soil. In addition, they increase the normal stress in the surrounding soil and compact the surrounding soil if it is cohesionless. When testing was com- plete on the compacted fill, 30-in. diameter geopiers were installed in a grid pattern south of Pile Cap 4. Plan and pro- file drawings are shown in Figure 3-40. The RAPs were spaced at 36 in. center to center (c-c) in the direction of load- ing and 40 in. c-c in the direction transverse to loading. The 13-pier configuration consisted of 4 piers next to the cap, 5 piers in the middle row, and 4 piers in the row farthest from the cap. The row farthest from the cap was installed first and the row closest to the cap was installed last. Each column extended to a depth of 12.5 ft below the top of the pile cap. Dynamic cone penetration tests were performed on three of the columns and penetration resistance exceeded 40 blows per 1.75-in. 45 Figure 3-38. Comparison of load-displacement curves for Pile Cap 4 with compacted sand extending 5 ft beyond the cap and Pile Cap 1 in native clay without soil adjacent to cap. Figure 3-39. Comparison of load-displacement curves for Pile Cap 4 with and without sand adjacent to the pile cap. 0 50 100 150 200 250 300 350 -0.2 0.0 0.2 0.4 0.6 0.8 1.0 1.2 1.4 1.6 Displacement (in) Lo ad (k ip s ) Compacted Sand Beyond Cap T3 Cap 4 Untreated Clay After Excavation T2 Cap 1 0 50 100 150 200 250 300 350 0 0.2 0.4 0.6 0.8 1 1.2 1.4 1.6 1.8 2 Displacement (in) Lo ad (k ip s) Sand Adjacent to Cap T4 Cap 4 No Sand Adjacent to Cap T3 Cap 4

Figure 3-40. Plan and profile view of Pile Cap 4 showing the locations of the rammed aggregate piers and location of excavated zone for subsequent test.

RAPs are a relatively inexpensive means of retrofitting a pile cap, but they are not designed specifically to increase lateral resistance. Nevertheless, comparison tests were performed to explore the potential for increasing lateral resistance using this approach. Test Results for Rammed Aggregate Piers Figure 3-41 provides a comparison of the lateral load- displacement curves for Cap 4 after treatment with RAP columns (Test 6) in comparison with the same cap without the columns (Test 5). At a displacement of about 1.3 in., the addi- tion of the RAP columns increased the total lateral resistance by about 40 kips. This represents an increase of about 15% rel- ative to the cap without the RAP columns. Figure 3-42 plots the load-displacement curves for Cap 4 after treatment with RAP columns before (Test 6) and after excavation (Test 7) of the soil adjacent to the pile cap. Because of reloading effects, the curves for Test 7 at small displacements are not particularly meaningful; however, at larger displacements they appear to be reasonable based on comparison with similar tests. The differ- ence between the curves for Tests 6 and 7 would represent the 47 Figure 3-41. Comparison of load-displacement curves for Pile Cap 4 with RAP extending to the top of the cap relative to tests on Cap 4 without RAP columns and Cap 1 in untreated native clay. Figure 3-42. Comparison of load-displacement curves for Pile Cap 4 after RAP treatment with and without soil adjacent to the pile cap. 0 50 100 150 200 250 300 350 -0.2 0 0.2 0.4 0.6 0.8 1 1.2 1.4 1.6 Displacement (in) Lo ad (k ip s) RAP to Cap Top T6 Cap 4 Untreated Clay to Cap Top T5 Cap 4 Untreated Clay T1 Cap 1 0 50 100 150 200 250 300 350 -0.4 -0.2 0 0.2 0.4 0.6 0.8 1 1.2 1.4 1.6 Displacement (in) Lo ad (k ip s) RAP to Cap Top T6 Cap 4 RAP to Cap Base T7 Cap 4 Untreated Excavated to Cap Base T2 Cap 1

passive force contributed by the clay with RAP columns. At a displacement of 1.25 in., the interpreted passive force is about 85 kips. This is about 35 kips higher than the passive force (50 kips) obtained when clay alone was acting against the pile cap. This result indicates that nearly all of the increased resistance provided by the RAP columns was a result of increased passive force against the pile cap and only about 10% of the increase was a result of increased soil-pile resistance below the pile cap. This result seems consistent based on the results of previous soil improvement tests in which little improve- ment in lateral pile resistance was achieved unless the soil improvement extended to the face of the piles. The increase of 35 kips in the passive force alone represents an increase of about 70% relative to the passive resistance provided by the clay alone. To produce a more readable report, additional plots, simi- lar to those presented for the pile group in virgin clay are not presented here but are available in Lemme (2010, Appendix 4). 3.11 Summary of Increased Resistance from Soil Improvement Methods and Cost Considerations A summary of the geometries of the various soil improve- ment techniques and the increase in lateral resistance that they produced is provided in Table 3-4. The greatest increase in lat- eral load (400 to 500 kips) was produced by the jet grouting method. Soil mixing and flowable fill also produced significant increases in lateral resistance (140 to 170 kips). Excavation and replacement techniques produced a relatively small increase in lateral resistance (20 to 40 kips). It should be noted that the treatments producing the great- est improvement were typically those that involved the larger volumes of treated soil and the greatest cost to implement; however, this is not always the case. To provide some indica- tion of the cost effectiveness of the various treatment methods relative to driving more piles, some simplified cost estimates were produced for the various approaches. A complete cost assessment is beyond the scope of this investigation and would be dependent on a number of factors that would vary from location to location. In addition, the geometries could poten- tially be further optimized to produce greater resistance rela- tive to the geometries used in the field tests. Furthermore, it may be possible to reduce the cost of ground improvement obtained by jet grouting by using lower strength mixes or by using more economical approaches such as soil mixing, as will be investigated subsequently. Nevertheless, this simple com- parison provides a first estimate of the economic viability of the ground improvement approaches for increasing lateral pile group resistance. One common alternative to soil improvement would be to simply add more piles and increase the size of the pile cap. According to the test results for Cap 1 during Test 2, the max- imum lateral load resisted by the nine-pile group was about 230 kips. Assuming this load is distributed evenly, each pile would have carried about 26 kips. Therefore, to obtain the same lateral resistance of 500 kips that was achieved through jet grouting beneath the pile cap, 20 piles would have to be added. Similar calculations have been made for each improve- ment approach. The cost of the additional piles, neglecting mobilization costs, were estimated by assuming a typical pile length of 80 ft, pipe pile costs of $30/ft, and driving costs of $12/ft. In addition, the cost of concrete fill and reinforcing steel cages in the piles was estimated assuming $150/cubic yard of concrete. Finally, adding more piles will require an increase in the size of the pile cap and an estimate of this cost also was made for each case assuming 3-ft center-to-center pile spacing. The cost of jet grouting was estimated to be roughly $475 per cubic yard for the 20% by weight cement content used in this study. Mobilization costs are highly variable and were not included in the cost because mobilization costs also were excluded from the pile driving costs. Soil mixing was assumed 48 Treatment Treatment Untreated Increase in Percent Treatment Method Dimensions Volume Resistance Resistance Increase Comments (LxWxD) (cu. Yds.) (kips) (kips) in Resistance Jet Grouting Below Cap 15'x10.5'x10' 58.3 282 500 160 Jet Grouting Adjacent to Cap 6.6'x13'x12' 38.1 214 398 185 Soil Mixing Adjacent to Cap 4'x11'x10' 16.3 282 170 60 Weak Flowable Fill Below Cap 13.5'x8.8'x6' 26.4 232 24 10 Flowable Fill Adjacent to Cap 6'x12'x6' 16.0 265 145 55 Compacted Fill to Edge of Cap 9.6'x8.75'x3.5' 10.9 232 23 10 Compacted Fill 5 ft beyond Edge of Cap 14.6'x8.75'x3.5' 16.6 232 40 18 Rammed Aggregate Piers Adjacent to Cap Top 13-2.5' dia x 13' deep 29.5 285 40 14 Rammed Aggregate Piers Adjacent to Cap Top 13-2.5' dia x 10.5' deep 23.6 50 35 70 1 Note 1: Increase in resistance is for passive resistance only Table 3-4. Summary of treatment geometries and increased resistance provided by the various soil improvement methods.

Table 3-5. Summary of increased resistance provided by the various soil improvement methods along with cost savings relative to providing additional piles. Increase in Equivalent Add'l Ground Savings Ground Treatment Method Resistance Number of Pile/Cap Improvement Relative to Percent Improvement Comments (kips) Pipe Piles Cost Cost Piles Savings Cost/kip Jet Grouting Below Cap 500 20 $84,200 $28,500 $55,700 66 $57 Jet Grouting Adjacent to Cap 398 16 $69,360 $38,000 $31,360 45 $95 1 Soil Mixing Adjacent to Cap 170 7 $30,345 $10,000 $20,345 67 $59 1 Weak Flowable Fill Below Cap 24 1 $4,335 $3,180 $1,155 27 $133 Flowable Fill Adjacent to Cap 145 6 1 $26,010 $3,600 $22,410 86 $25 1 Compacted Fill to Edge of Cap 23 $4,335 $544 $3,791 87 $24 Compacted Fill 5 ft beyond Edge of Cap 40 2 $8,670 $828 $7,842 90 $21 Rammed Aggregate Piers Adjacent to Cap Top 40 2 $8,670 $4,225 $4,445 51 $106 Rammed Aggregate Piers Adjacent to Cap Top 35 2 $8,670 $4,225 $4,445 51 $121 2 Note 1: Cost of soil improvement doubled to account for increased resistance in opposite direction Note 2: Increase in resistance is for passive resistance only Note 3: Cost/kip for pile/pile cap = $182/kip

50 provided cost savings, although in some cases, the cost savings are small, as is the increase in resistance. This is particularly true for the excavation and replacement approaches. Finally, the cost per kip of increased lateral resistance was computed for each case to provide another indication of the cost- effectiveness of the various approaches. For comparison pur- poses, the cost per kip for the pipe pile alternative was $182/kip. All of the ground improvement methods had lower costs per kips than that for the piles. The lowest cost per kips was pro- vided by the excavation and replacement method because of the low cost of the treatment method. However, it must be rec- ognized that despite the low cost per kip for this treatment method, the potential for increasing resistance also was quite limited. to cost about $300 per cubic yard for the 10% by weight cement content, and flowable fill was assumed to cost $75 per cubic yard in addition to excavation costs. For the treatments adja- cent to the pile cap, the cost was doubled assuming that a sim- ilar improvement zone would be required on the opposite side to account for load in the opposite direction. The excavation and replacement cost was assumed to be $50 per cubic yard for the small volumes involved and the cost of RAPs was assumed to be $50 per ft of length. The estimates of improvement costs associated with each treatment method are shown in Table 3-5. In addition, the cost savings associated with soil improvement compared to providing additional piles is listed for each case along with the percent savings relative to additional piles. In all cases, the ground improvement method

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Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils Get This Book
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TRB’s National Cooperative Highway Research Program (NCHRP) Report 697: Design Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils examines guidance for strengthening of soils to resist lateral forces on bridge pile foundations.

The report presents computational methods for assessing soil-strengthening options using finite-element analysis of single piles and pile groups and a simplified approach employing commercially available software.

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