Recommended AASHTO Guide Specifications for ABC Design and Construction (2018)

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NCHRP Project 12-102 58 C H A P T E R 5 ABC Design Specification Development 5.1 Approach This project is essentially a large-scale synthesis of past work. No formal laboratory research was completed under this project. The goal was to find past research, review the results and then measure the readiness of the technology for deployment into the AASHTO community. There were several steps in the development of each section: 1. The team reviewed past projects that have been completed in each state. The goal being to include as many ABC technologies as possible. The number of technologies are numerous, covering virtually all elements in a bridge. From this listing, pertinent research was grouped for the development of provisions for this guide specification. For technologies that did not have significant research, more general provisions were developed. 2. Pertinent research was reviewed and compared. This was not always easy to accomplish due to the different approaches used by each past research team. Some of the data was not completely in sync. Where possible, the results of multiple research projects were compared. The results were also tested against the proposed provisions to ensure that the provisions are applicable and conservative. Research results were also compared with other structural codes and documents including the ACI Building Code and Commentary, and manufacturer literature. 3. The results of the research comparison led to the development of proposed provisions. These provisions were compared with current AASHTO LRFD provisions to determine if they could be referenced. If no AASHTO provisions exist, new provisions were developed. 4. A technology readiness evaluation was undertaken for each provision. The following measures were used for the technology readiness evaluation: a. Testing and Research: Level of study performed to date b. Existing Specifications: Availability of data or other code provisions that may be applicable c. Implementation: How often the technology has been used in the past d. Durability: Performance history of the technology These measures are scored on a scale of 1 to 5 and then weighted according to importance of the measure. The weight factors are somewhat subjective. The following weight factors were developed by the project research team and approved by the project panel: a. Testing and Research: The second highest factor of 25 was applied to this measure. This underscores the need for solid research to back up a specification provision. b. Existing Specifications: A relatively low factor of 15 was used for this measure, since most of these technologies do not have existing specifications.

NCHRP Project 12-102 59 c. Implementation: A high factor of 30 was applied to this measure. The thought was that real world experience would be the best test of a technology. d. Durability: A high factor of 30 was applied to this measure based on the owner’s response to the questionnaire where durability was noted to be a significant concern. Figure 5.1-1 is a sample technology readiness evaluation form. Figure 5.1-1 Sample Technology Readiness Evaluation Form Each technology-related specification section was evaluated using this method. Only technologies with a score of 75 and higher were included in the sample specification. Several technologies scored very close to 75 and were included. Justifications for these inclusions are in the applicable following sections.

NCHRP Project 12-102 60 5.2 Specification Section and Article Development The team was charged with developing guide specifications in AASHTO format. This is not a stand- alone design specification, but a supplement to the AASHTO LRFD Bridge Design Specifications. The key features of AASHTO Guide Specifications include: 1. Two-column format: Specifications in the left column, commentary in the right column. 2. A “section” is akin to a chapter in other documents. 3. Articles are numbered using number headings (1.1, 1.1.1, 1.1.1.1, etc.) 4. Commentary headings start with a Capital C. 5. Tables of Contents are generated for each section as opposed to an overall document Table of Contents. 6. The first article in each section contains the scope of that section. 7. Notations for each section are included in the front of each section. Reference to the applicable articles is also identified. 8. References to current AASHTO provisions do not include the actual provision number, just the provision title. This is done to address possible re-organization of AASHTO documents. There is a hierarchy in the use of AASHTO documents. Standard AASHTO specifications are mandatory specifications (unless overridden by agency specifications). Guide Specifications are different, but similar to standard AASHTO specifications. They differ in that they are not mandatory. Guide specifications are written similar to a full specification. The guide specifications make use of the terms “shall”, “should”, “may”, and “recommended”. The following describes how these terms should be interpreted: • The term “shall” denotes a requirement for compliance with the specifications. • The term “should” indicates a strong preference for a given criterion. • The term “may” indicates a criterion that is usable, but other local and suitably documented, verified, and approved criterion may also be used in a manner consistent with the LRFD approach to bridge design. • The term “recommended” is used to give guidance based on past experiences. The completed Guide Specification for ABC is not included in this report, since it will be published by AASHTO in the near future. The following sections contain information on the development of each specification section. Where appropriate, a technology readiness evaluation was completed and is contained herein. 5.3 Section 1 - Introduction The provisions in this section serve as an introductory section of the overall guide specification. Section 1 includes definitions of common prefabricated elements and systems, recommended roles and responsibilities for ABC projects and design considerations for different types of ABC projects. The following are brief explanations of the content of each major article and the information used in the development of the article. 5.3.1 Article 1.3 - Definitions The definitions were vetted through the project panel and the AASHTO T-4 Construction Technical Committee. These definitions are also used in national project databases. Use of common definitions aids in the management and dissemination of information across the country. The definitions are organized by major category for ease of use. The definitions are included in Chapter 1 of this report.

NCHRP Project 12-102 61 5.3.2 Article 1.4 - Design Responsibilities for Prefabricated Elements A review of project histories across the country had revealed some confusion regarding design responsibilities for projects with prefabricated elements. The confusion arises from other industries. In the precast parking facility market, the final design of the elements is left to the fabricator. This is due to the repetitive nature of parking structure design. Bridges are a different situation. Each bridge requires specific design requirements that should be completed by the design engineer. The roles of the designer are defined in the article. Even if an element is designed and detailed by the designer, there is still a design role to be played by the contractor (including fabricator). In most cases, the design for shipping and handling of elements falls under the purview of “contractor means and methods”. The provisions in this section were written to be consistent with practice in most states and most project constructed to date. 5.3.3 Article 1.5 - Design Responsibilities for SPMT Systems SPMT Systems require significant engineering design on the part of the contractor; however the design still carries a significant amount of design responsibility. The designer needs to verify that the bridge system can be lifted and moved without damage based on an assumed lifting method. In order to accomplish this, the designer needs to make certain assumptions, including preliminary SPMT layout and methods of supporting the bridge in the staging area. The provisions of this article includes recommendations for these assumptions. The role of the contractor is also included in this article, which primarily includes the design of the SPMT units and falsework. Several key documents were used as the basis of these provisions including the Utah DOT Structures Design and Detailing Manual (2015) and the FHWA Manual on Use of SPMT to Remove and Replace Bridges (2007). The experience of the author’s firm on several SPMT installations was also used for the development of this article. 5.3.4 Article 1.6 – Design Responsibilities for Lateral Slide Systems The provisions for this article are similar to the provisions for SPMT Systems. The designer and the contractor both play a role in the execution of the design and construction of the project. The Utah DOT Structures Design and Detailing Manual (2015) was used as the basis for this article. 5.4 Section 2 - General Design Provisions This section contains information on design parameters including loads and load factors for prefabricated elements and bridge systems. 5.4.1 Article 2.4 – Loads and Load Combinations This article contains recommendations for loads and load combinations for specific ABC methods. The design of elements within a bridge are governed by loads and load combinations in the AASHTO LRFD Bridge Design Specifications. Items that are not covered in existing specifications include load factors for shipping and handling of elements, and load and load factors for bridge systems.

NCHRP Project 12-102 62 No research was discovered regarding loads in element during fabrication, shipping and erection. The Precast/Prestressed Concrete Institute Design Handbook MNL-120 (2010) does contain recommended load factors for these processes. This handbook has been used by fabricators for years with good results, therefore it is the opinion that it is sufficient for this guide specification. There are no published reports or documents covering loads and load factors for bridge systems. There have been studies by the Utah DOT regarding the stresses in bridges during SPMT moves (Rosvall, 2010). While this information is interesting, it is limited to several bridges and one type of falsework used during the moves. Fortunately, NCHRP has a specific research project that is studying dynamic effects of bridge systems. Project 12-98 investigated the dynamic effects of bridge system moves, including the effects of falsework and bridge stiffness. Project 12-98 was undertaken in a parallel timeframe with this project. Michael Culmo was the principal investigator of both projects, therefore the results of Project 12-98 were included in this article. Reference is made to one of the deliverables for Project 12-98, a document entitled Guidelines for Dynamic Effects for Bridge Systems. 5.5 Section 3 - Design of Prefabricated Elements This section is the largest section (78 pages) in the Design Guide Specification. It covers the design of elements, connections, geosynthetic reinforced soil/integrated bridge system, and accelerated backfilling. Many of the articles in this section are based on recent research. Where applicable, the research was referenced and a technology readiness evaluation completed. 5.5.1 Article 3.4 – Seismic Design for ABC The requirements included in this article and the remainder of this section are supplementary to, rather than a substitute for, those in the AASHTO Guide Specifications for LRFD Seismic Bridge Design (AASHTO SGS), and the AASHTO LRFD Bridge Design Specifications (AASHTO LRFD). The three documents should therefore be used in tandem. There are aspects of AASHTO LRFD that are not compatible with ABC connections, therefore limitations are provided in the proposed ABC Guide Specifications. These incompatibilities are generally applied when an evaluation of a components performance is needed to validate proper force transfer, capacity development, rotational capacity, or other behavior that could compromise the intended seismic performance. Several features of the AASHTO SGS are adopted in these proposed requirements. One is the goal of achieving a structure that suffers only minimal damage during moderate earthquakes and does not collapse during rare earthquakes. A second is the need to define an Earthquake Resisting System and corresponding Earthquake Resisting Elements for zones of moderate to high seismic risk. A third is the adoption of a Type I design strategy, according to which any inelastic behavior should occur in a ductile mechanism, such as bending, and should be restricted to the substructure, while the superstructure and foundations remain elastic. This behavior can be ensured through the use of capacity design principles. The proposed specifications are intended to assist the designer in establishing reliable and continuous load paths over which to transfer loads from their point of origin through the appropriate elements and into the supporting soil. In many cases those elements will be prefabricated in a plant, and joined by connections that are completed in the field. Connections are a critical part of any load path, particularly for structures subjected to seismic loading, so they require close attention in design. Many of the systems

NCHRP Project 12-102 63 addressed by the proposed specifications contain connections, and provisions for their design. Those provisions are based on laboratory testing and computational modeling. References permit the designer to review the source documents for background information. A distinction is made between a joint and a connection. A joint is the region where two members intersect, while a connection is the region where two previously separate elements are connected structurally so they act together. Connections are often, but not always, made at joints, largely because that strategy permits the convenience of transporting straight elements to the site. As a counter-example, it would be possible, but probably not convenient, to build cruciform members in a plant and connect them on site at mid-span. Then, the connections would not be made at the joints. In these proposed specifications, the focus is on the performance of the connection. This does not relieve the designer of the responsibility for verifying the integrity of the load path through the remainder of the joint. In some cases the properties of even a ductile connection may affect the system response. This may occur, for example, when grouted connectors are used. The sleeve is likely to be relatively rigid, and, if it is placed in the column, it may inhibit plastic hinging there. That will force the inelastic action to occur elsewhere and may reduce the ductility capacity of the overall system. In those cases a reduced response modification factor is specified for the AASHTO LRFD forced-based design and a reduced plastic hinge length is specified for AASHTO SGS displacement-based design. Both approaches and the associated equations are justified by analytical and experimental research. 5.5.2 Article 3.5 – Prefabricated Element Design The provisions in this article reinforce the concept that prefabricated elements need to be design for all forces and using all load combinations specified in the AASHTO LRFD Bridge Design Specifications. In the case of precast concrete elements, the most common approach is emulative design, where an element and the connections between elements are detailed to emulate a similar cast-in-place concrete element. In this case, the design of the elements is the same as if the bridge were to be built using conventional construction. The AASHTO LRFD Bridge Design Specification is applicable and appropriate for this situation. 5.5.3 Article 3.6 – Connection Design and Detailing There are many different types of connections that are used for prefabricated elements. This article has been divided into several sub-articles that address the different connections that are in use. 5.5.3.1 Article 3.6.2 Cast-in-Place Concrete Closure Joints Using Lapped Bar Reinforcement This sub-article contains information on some the most common connections in use. Closure joints (also known as closure pours in some jurisdictions) made with lapped reinforcing bars are popular, primarily because they are an emulative connection that can be readily understood by designers. Text was included in this article to accommodate the approach to using initial strength design of closure joint connections. In this scenario, the connection is designed for a concrete strength that is less than the specified concrete strength. Closure joints are problematic in very fast construction projects because typical concrete products do not gain strength fast enough to accommodate the accelerated schedule. The use of initial strength design allows for reduced construction time without sacrificing durability or quality.

NCHRP Project 12-102 64 An example of this was used on the 93Fast14 Project in Massachusetts. The closure pours were specified to have 4000 psi concrete based on typical concrete used in bridge decks in Massachusetts. The deck connections were designed for 2500 psi concrete, but the material specification was for a 28 day compressive strength of 4000 psi. This approach resulted in a slightly decreased bar spacing, but allowed for the bridge to be opened to traffic much faster. The nature of concrete strength gain is that most concretes can gain 2500 psi relatively quickly. The additional 1500 psi takes longer to achieve. This approach is commonly used in the production of prestressed concrete girders. Text was included to inform designers that they should account for tolerances in closure joint reinforcement detailing. This is more of a construction issue; therefore, the text was kept brief. The project team has seen a number of projects where this was ignored, which resulted in problems. Fortunately, NCHRP has a specific research project (Project 12-98) that is studying tolerances in prefabricated elements and systems. Project 12-98 was undertaken in a parallel timeframe with this project. Michael Culmo was the principal investigator of both projects, therefore the results of Project 12- 98 were included in this article. Reference is made to one of the deliverables for Project 12-98, a document entitled Guidelines for Prefabricated Bridge Elements and Systems Tolerances. 5.5.3.2 Article 3.6.2.1 Reinforcing Bars This article contains design information for various types of lapped reinforcing bar connections that are in use. The use of straight lapped bars is covered in the AASHTO LRFD Bridge Design Specifications, therefore a reference is called for. The following is a technology readiness form for this article:

NCHRP Project 12-102 65 Figure 5.5.3.2-1 Technology Readiness Evaluation for Article 3.6.2.1

NCHRP Project 12-102 66 5.5.3.3 Article 3.6.2.2 Hooked Reinforcing Bars The current AASHTO LRFD Bridge Design Specifications for Hooked bars do not specifically mention lap splices of hooked bars. The tension lap splice specification refers to tension development length ld, but not hook development ldh. The project team reviewed the history of the lap splice provisions and determined that the adjustment factors for lap lengths are only applicable to straight bar laps. The failure mechanism for hooked bars is quite different than straight bars. Hooked bars fail by a combination of development along the straight portion of the hook and crushing of the concrete at the bend. Several studies were identified during the literature search for this project. The most significant recent study was NCHRP Project 10-71 (French, et al., 2011). The following is a brief overview of what was found in that project report along with the synthesis of other similar work: 1. NCHRP Project 10-71 Cast-in-Place Concrete Connections for Precast Deck Systems (French, et al., 2011): The project studied several potential connections for deck panels and decked beam systems. Lapped, hooked bars were chosen as a detail with high potential due to cost, performance and ease of construction. The recommendations of this project were to use the ACI Hook development length equation for the lap length. It should be noted that the AASHTO Hook Development length equation gives the same results, but with the terms adjusted to reflect different units (psi vs. ksi). The AASHTO equation is further simplified by setting fy equal to 60,000 psi. 2. Connection of modular steel beam precast slab units with Cast-in-place closure pour slabs (Brush, 2004): This research at Texas A&M University studied the strength of hooked bars that were lap spliced together. Several details were studied including longitudinal joints between decked beam elements and closure pours at link slabs. Each test used hooked bars that were set up as contact lap splices. The length of the laps was equal to or better than the hook development length specified in the AASHTO LRFD Bridge Design Specifications. No transverse bars were included in the testing. The results of the tests indicated that the lapped hooked bars with very small lap lengths had flexural capacity that was slightly less than the full strength capacity of the bars. Hooked lap splices with longer lap lengths were found to be adequate. The lack of transverse bars within the hooked bars may explain the slight loss of strength for shorter laps. 3. Precast Alternative for Flat Slab Bridges (Sheng, et al., 2013): This research project investigated the use of prestressed double tee beams for short-span bridges. The work included an extensive investigation of the connection between the double tee beams. This connection is primarily a small closure pour with lapped reinforcing bars. Several of the tests included lapped hooked bars. Various grouts and UHPC were tested. The details used #4 hooked bars with a 6” lap length. Two transverse #4 bars were also included inside the hook. All of the grouts tested contained some measure of fibers in the mix. Both steel fibers and polyvinyl alcohol (PVA) fibers were tested. The steel fibers were noted to be structural fibers, while the PVA fibers were noted to be used more for architectural reasons (crack control). The use of fibers in the non-UHPC grouts makes direct comparison of results to other research difficult. In general, the results were consistent with the results of NCHRP Project 10-71 (French, et al., 2011). The lapped hooked bars were capable of developing the bars within the AASHTO LRFD specified development length for the hook. The testing completed under NCHRP Project 10-71 (French, et al, 2011) used very close bar spacing (4.5”) and transverse bars under the bearing zone of the hook. Both of these factors should improve the performance of the connection. This presumption is due to the fact that the failure mechanism of a

NCHRP Project 12-102 67 hooked bar is a combination of concrete bond on the bar combined with crushing of the concrete under the hook. The transverse bars serve two purposes. They increase the bearing area of the hook and improve the force transfer through the failure cone via shear friction. By examining all of the relevant research projects, it appears that transverse bars are important for details with very short lap lengths. The transfer mechanism of forces in non-contact lapped hooked bars is a concern. The theoretical traditional failure mechanism is crushing of the concrete under the hook which is similar to a headed anchor road under direct tension. The failure plane would be a cone surface. Non-contact lapped hooked bars that are spaced too far apart would end up with non-intersecting failure cones, which would limit the amount of force transfer. Several ways to approach this problem were investigated based on the information obtained in the literature search. One approach could be to limit the maximum spacing of the hooks in a similar fashion as a non-contact lap splice. Another approach could be to adjust the lap length based on the non-contact bar spacing. The NCHRP Project 10-71 (French, et al, 2011) research included a study of hooked transverse bars. The performance of these test specimens indicate that a standard development length can be used as the lap length as long as transverse bars are incorporated into the connection. The transverse bars used for the testing were of equal size to the hooked bars. For this reason, the project team recommends matching the size of the transverse bar to the hooked bar. This is the basis for the proposed provisions. Even with the use of transverse bars, the team recommends that there be a limit on the spacing of non- contact lap splices. The maximum lap bar spacing provision for non-contact straight bar lap splices seems appropriate based on the above discussion; therefore, it was also added. A value of 4” is recommended and is consistent with the research testing of these connections. The following is a technology readiness form for this article:

NCHRP Project 12-102 68 Figure 5.5.3.3-1 Technology Readiness Evaluation for Article 3.6.2.2

NCHRP Project 12-102 69 5.5.3.4 Article 3.6.2.3 Headed and Mechanically Anchored Deformed Reinforcing Bars NCHRP Project 10-71 (French, et al, 2011) studied the use of headed reinforcing bars in closure joints. The researchers studied the use of Number 5 headed bars with a 6” lap length. These bars come in two head sizes. The smaller head size was chosen for the study. The concrete strength chosen for the testing was 6000 psi. The testing showed that the bars could be fully developed with the 6” lap and center to center spacing of 6” (3” non-contact lap splice). The NCHRP Project 10-71 (French, et al, 2011) research was somewhat limited. The number of configurations was not sufficient to develop a design specification equation. The parameters that were not varied including the size and spacing of the bars, the head diameter, the concrete strength and the spacing. NCHRP Project 12-69 (Oesterle, et al., 2009) also investigated cast-in-place concrete closure joints for connecting deck bulb tee girders. This research investigated several different connection details during the initial portions of the work. The recommended connection type involved the use of a single row of headed lapped reinforcing bars. As with NCHRP Project 10-71 (French, et al, 2011), the number of tests and the test parameters were limited. The only variables were lap length (2.5”, 4” and 6”), bar spacing (4” and 6”) and bar type (welded wire fabric and headed bars). This only represents a limited amount of variation. The work did not include other variables such as concrete strength (7000 psi), bar coating (uncoated bars), and bar sizes (#5). The test results proved that a #5 headed bar could be lap spliced in 6” with bars spaced 6” on center (3” non-contact lap splice). Two transverse bars were placed within the splice. They were placed mid-way along the splice length. The tests with the smaller lap lengths and welded wire fabric did not produce adequate strength, ductility and rotational capacity. The results of these two studies of lapped headed reinforcing bars in a concrete closure pour realistically represents only several data points. The bar size and spacing was the same in both projects. The only variable was two concrete strengths. This amount of data is not sufficient to write design specifications; therefore, the team investigated other potential sources of design guidance. The AASHTO LRFD Bridge Design Specifications do not contain provisions for headed reinforcing bars; however, there are other specifications that do cover headed bars (see discussion below). a. ACI 318 Building Code Requirements for Structural Concrete: The ACI Building code does contain specifications for development length of headed bars. These provisions include adjustments for epoxy-coated bars and concrete strength; however, there are no provisions for lap splices. a. HRC: This manufacturer has had an independent product evaluation report prepared, which is included in their product information booklet. This report includes recommended provisions for calculating development length and lap splice lengths for headed bars that takes into account the bar spacing. In order to gain an understanding of these various recommendations, a comparison was made between ACI-318, NCHRP Project 10-71 (French, et al, 2011), NCHRP Project 12-69 (Oesterle, et al., 2009), and recommendations from HRC. The results of the NCHRP Project 10-71 (French, et al., 2011) tests and the HRC recommendations were plotted against the ACI provisions In order to see if the ACI provisions are appropriate. Figure 5.5.3.4-1 shows the results of this comparison.

NCHRP Project 12-102 70 Source of data: (French, et al, 2011)), 12-69 (Oesterle, et al (2009)), and Recommendations from HRC Figure 5.5.3.4-1 Comparison of ACI 318 Article 12.6 with NCHRP Projects 10-71 0.00 5.00 10.00 15.00 20.00 25.00 3 4 5 6 7 8 9 10 11 De ve lo pm en t L en gt h Bar Size ACI 318-08 Article 12.6 f'c = 9500 psi = NCHRP 10-71 Mix ACI-318 Article 12.6 HRC Recommended Lap Length HRC Recommended Dev. Length NCHRP 10-71 Results 0.00 5.00 10.00 15.00 20.00 25.00 30.00 3 4 5 6 7 8 9 10 11 De ve lo pm en t L en gt h Bar Size ACI 318-08 Article 12.6 f'c = 7000 psi = NCHRP 12-69 Mix ACI-318 Article 12.6 HRC Recommended Lap Length HRC Recommended Dev. Length NCHRP 12-69 Results

NCHRP Project 12-102 71 The concrete strengths used for these comparisons are the approximate concrete strengths used in the research. The concrete strength for NCHRP 10-71 (French, et al, 2011) tests was approximately 9400 psi. The concrete strength for NCHRP 12-69 (Oesterle, et al. 2009) tests was approximately 7000 psi. The few tests of the headed bars in Project 10-71 (French, et al., 2011) and Project 12-69 (Oesterle, et al. 2009) fall very close to the values calculated by the ACI-318 Provisions. The HRC recommended development equation is overly simplistic (8 bar diameters or 6 inches); therefore, it is not recommended. The HRC recommended lap lengths are above the ACI-318 values due to the presence of a 1.3 safety factor in their equation. There are two potential approaches for a recommended guide specification provision. 1. Use ACI-318 Section 12.6 along with the HRC Adjustment factor equation: This approach would require a verification of the 1.3 factor in the HRC equation. 2. Set the lap length equal to the ACI-318 Section 12.6 value, require transverse bars, and limiting the maximum spacing of bars. Approach 1 is not recommended. The 1.3 factor appears to be based on limited studies of this one bar manufacturer’s product. Design recommendations for lap splices from other manufacturers were investigated, but none were found. This may be due to the fact that these products are commonly used to connect beams to columns in buildings, where the headed bar is anchored into mass reinforced concrete, not as a lapped detail. Research completed under NCHRP Project 10-71 (French, et al., 2011) and Project 12-69 (Oesterle, et al., 2009) did include transverse bars and closely spaced headed bars. The maximum spacing between lapped bars was 3” in these tests. The results of the few tests completed indicate that the lap length be assumed to be equal to or slightly shorter than the ACI-318 specified development length. As with hooked bars, the failure mechanism of a headed bar is a combination of bar bond and crushing under the bearing portion of the bar (head). The addition of transverse bars within the tension failure cone of the bar would in theory, increase the capacity due to shear friction behavior as the transverse bar passes through the failure cone. Placement of transverse bars directly under the head will also increase the crushing zone resistance (in theory). The project team recommends approach 2, which is consistent with the work completed under NCHRP Project 10-71 (French, et al., 2011) and Project 12-69 (Oesterle, et al., 2009). The provisions presented in the article are based on ACI-318 Article 12.6 and ACI-318 Article 3.5.9 with the addition of a requirement to have transverse bars placed under the heads of the bar and the maximum spacing set at 3” for non-contact lap splices. The following is a technology readiness form for this article:

NCHRP Project 12-102 72 Figure 5.5.3.4-2 Technology Readiness Evaluation for Article 3.6.2.3

NCHRP Project 12-102 73 5.5.3.5 Article 3.6.2.4 Reinforced UHPC Connections There have been a significant number of recent research projects that investigated the use of reinforced UHPC for connections of prefabricated elements. In the opinion of the project team, the FHWA Tech Note entitled Design and Construction of Field-Cast UHPC Connections (Graybeal, 2010) represents the current a state of practice in the use of UHPC for prefabricated bridge element connections. The document is a synthesis of past work by the FHWA Research Laboratory and others. The document contains recommended design and construction specification language. The provisions of this document were compared with the other research project findings to verify that the provisions are consistent with test results. The following is a synopsis of this comparison: 1. Design Guide for Precast UHPC Waffle Deck Panel System, including connections (Aaleti, et al., 2013): This guide was developed by FHWA for a research project involving UHPC precast deck elements connected with UHPC joints. A recommendations contained in the report states, “The reinforcement, including strands, shall be designed with a minimum clear spacing equal to 3.25 times the diameter of the reinforcement or 1.5 inches, whichever is greater.” This recommendation has been reduced in the FHWA Tech Note document. Based on this, the more conservative values of 1.5 times the fiber length is proposed. The panel to panel connection tested was specific to the waffle slab system. The reinforcement was a 1” diameter stainless steel deformed dowel lapped approximately 7”. The testing showed that it would function well; however, design criteria for this connection was not developed. The value of 7” is close to the FHWA Tech Note document recommended value of 8 bar diameters. 2. SHRP2-R04 Innovative Bridge Designs for Rapid Renewal (2014): This project included the investigation of UHPC connections for decks (both transverse and longitudinal). A joint detail with lapped hooked bars was chosen for testing. The detail contained #5 bars with a 6” lap splice. Four transverse bars were also placed within the hooks. The results of the testing indicated that the detail provided “more than adequate” strength. Fatigue testing was also completed. No issues were noted. There were issues with the bond of the UHPC with the adjacent precast deck element. The requirements suggested by the FHWA Tech Note document for roughened and exposed aggregate surface should address this problem. 3. Feasibility Analysis of UHPC for Prestressed Concrete Bridge Applications (Weldon, et al., 2012): This project focused on the use of UHPC for girders. Connection details and design were not included in the work. 4. High Performance Joints for Concrete Bridge Applications (Harryson, 2003): This research project from Sweden included the investigation of Compact Reinforced Composite (CRC) Concrete. This material is very similar to UHPC with similar compressive strengths. The CRC mix did have a higher fiber content than US UHPC (6% by volume). The following are the results of the testing: a. The specimen tested had straight lapped reinforcing bars. The lap length was 100 mm (4”) for bars with a diameter of 16 mm (0.63”) and a yield strength of 564 MPa (82 ksi). Various cover

NCHRP Project 12-102 74 conditions were tested as well as different transverse bar arrangements. Fatigue testing was also completed. b. When compared to the recommended FHWA Guide (Graybeal, 2010), the tested configuration had a slightly undersized lap splice. If the tested detail was designed using the FHWA guide (Graybeal, 2014), the lap length would need to be 4.7”. The results of this research indicate that transverse bars were necessary in order to make the 4” lap work for the tested bar. A 4” lap detail without transverse bars failed prematurely (bond failure). The testing did not include longer lap lengths; therefore, it is difficult to know what lap length would work without the transverse bars. 5. Behavior of Field-Cast UHPC Bridge Deck Connections under Cyclic and Static Structural Loading (Graybeal, 2010): This research appears to be the basis for the development of the FHWA Tech Note (Graybeal, 2014). A variety of connection details were tested including headed bars, hooked bars and straight bars. The testing included static and fatigue testing. Water ponding was incorporated into the fatigue testing to study the long-term durability of the UHPC connections. 6. Behavior of UHPC Connections between Precast Bridge Deck Elements (Graybeal, 2010): This paper is synopsis of a portion of the previous paper discussed in item 6 above. 7. Precast Alternative for Flat Slab Bridges (Sheng, et al., 2013): This research project investigated the use of prestressed double tee beams for short-span bridges. The work included an extensive investigation of the connection between the double tee beams. This connection is primarily a small closure pour with lapped reinforcing bars. Various grouts and UHPC were tested. The details used #4 bars hooked bars with a 6” lap length. Two transverse #4 bars were also included inside the hook. All of the grouts tested were formulated with some measure of fibers in the mix. Both steel fibers and polyvinyl alcohol (PVA) fibers were tested. The use of fibers in the non-UHPC grouts makes direct comparison of results to other research difficult, since it is inherently different material. The results of the tests with typical UHPC with 2% steel fibers was consistent with the FHWA Tech Note. The synthesis of the past research work on UHPC connections in closure pours with lapped reinforcing bars indicates that the FHWA Tech Note (Graybeal, 2014) is consistent and applicable with all of the past work completed. This guide offers specific design specification guidance; therefore, the NCHRP Project 12-102 team recommends using the FHWA Tech Note (Graybeal, 2014) as the basis for the ABC Guide Specification section on UHPC connections. The organization and wording of the guide was modified to be consistent with other similar AASHTO Design Specifications. The following is a technology readiness form for this article:

NCHRP Project 12-102 75 Figure 5.5.3.5-1 Technology Readiness Evaluation for Article 3.6.2.4

NCHRP Project 12-102 76 5.5.3.6 Article 3.6.4.1 General In this section, connections between elements are made by connecting tension bars with mechanical reinforcing bar connectors. Those connectors allow tension to be transferred from one bar to another. In most cases the connected bars are collinear. Such connectors have been in use in practice since the seventies (Yee, 1973). However, two grades of mechanical connector are available. They are referred to in ACI 318-14 as Type 1 and Type 2. They are distinguished by the fact that Type 2 connectors must satisfy all the requirements of the Type 1 devices, but must, in addition, be able to develop the nominal tensile strength of the bars being connected, a requirement that Type 1 connectors cannot meet. ACI does not allow Type 1 connectors to be used within two times the member depth from the face of the connecting element. That definition encompasses, with some margin of safety, the potential plastic hinge region, and implies that Type 1 connectors are essentially to be used only in regions where yielding will not occur, to ensure that premature failure by bar pullout or coupler fracture do not compromise the system’s ductility capacity. ACI does allow Type 2 connectors to be used anywhere except within half a member depth of the face of the connecting member in special moment frames with ductile connections, unless the connections are “strong”, and yielding is forced to occur elsewhere, then the restriction does not apply there either. The AASHTO LRFD and SGS do not use Type 1 and 2 terminology. The SGS requires that, in SDCs C and D, any mechanical connector be placed outside the plastic hinge region, the length of which is defined by Priestley et al., 1996. However, in piles or shafts, where such placement is virtually impossible, mechanical connectors are allowed within the plastic hinge zone with the owner’s approval and provided that they can develop the tensile strength of the bar, i.e. they satisfy ACI’s Type 2 requirements. AASHTO LRFD is similar for Seismic Zones 3 and 4, but with limitations based on prescriptive criteria. In these proposed specifications, ACI’s distinction between Type 1 and 2 couplers is adopted, with only Type 2 connectors allowed within the plastic hinge region. Many tests have been conducted on columns connected to a foundation or cap beam by mechanical bar connectors, as well as on isolated couplers in air. Loadings have included static, fatigue, cyclic, and even high strain reversal simulating the effects of accidental explosions (Rowell et al., 2009). While strength has, for many years, been the primary criterion for their use, achieving sufficient ductility has recently emerged as a potential problem (Marsh et al., 2011). In many cases, the connection is made with a grouted steel sleeve, which is so rigid that it allows essentially no yielding within is length, and the entire elongation must take place in the bar adjacent to the end of the sleeve. If the sleeve is placed in the column, the bond of the bar in the foundation or cap beam is likely to be good, so the elongation must occur within a short distance at the interface, causing high local strains. Deliberately debonding the bar has been shown to improve ductility, but the matter is not yet fully resolved. Figure 1-1 shows the Technology Readiness Form for this article. Abundant research is available on the axial response of mechanical connectors as well as on the seismic behavior of column-pier cap or column- footing connections with mechanical splices. In addition, for precast moment resisting frames of buildings, mechanical connectors have also been used extensively in the field and so this item receives a high score for technology readiness. The following is a technology readiness form for this article:

NCHRP Project 12-102 77 Figure 5.5.3.6-1 Technology Readiness Evaluation for Article 3.6.4.1

NCHRP Project 12-102 78 5.5.3.7 Article 3.6.4.2 Type 1 Mechanical Connectors The definition of Type 1 connector presented in this article corresponds to a “full mechanical connection” as included in AASHTO LRFD. In the proposed guide specifications, Type 1 mechanical connectors are not allowed in plastic hinge regions for all SDCs or seismic zones. Experimental data exists from research studies and manufacturer’s test on the strength and ductility characteristics of these mechanical devices. Examples include: grouted splice sleeves (NMB) tested under monotonic loading by Elinea et al., 1995 and Jansson at Michigan DOT, 2008; Dynamic and impact testing of grouted splices by Noureddine, 1995 and Rowell and Hagger, 2010; Static and dynamic test of grouted sleeves and headed bars sleeves at the University of Nevada, Reno (Haber et al., 2013) and Tazarv and Saiidi, 2014); Multiple test of splice sleeves in Japan and Malaysia (NMB Splice Sleeve). AASHTO refers to these as full mechanical connectors (Type 1 terminology is from ACI 318-14). Because Type 1 mechanical connectors are already included in AASHTO design documents, and also because of the extensive research (Janssson, 2008; Noureddine et al., 1996; Rowell et al., 2009) and field applications with mechanical connectors this article scores high in technology readiness. The following is a technology readiness form for this article:

NCHRP Project 12-102 79 Figure 5.5.3.7-1 Technology Readiness Evaluation for Article 3.6.4.2

NCHRP Project 12-102 80 5.5.3.8 Article 3.6.4.3 Type 2 Mechanical Connectors The proposed Article 3.6.4.3 is a compromise between the ACI and AASHTO approaches, because it allows the use of mechanical splices at and near column-footing or column-pier cap interface but, for high seismic zones, it makes provision for the reduced displacement ductility capacity that has been observed in the laboratory with the use of mechanical connectors in plastic hinge zones. For the AASHTO LRFD force-base design in high seismic risk, the reduced ductility is handled in Article 3.6.4.4.1 through the specification of a reduced modification factor. On the other hand, the reduced ductility associated with using mechanical connectors in the SGS displacement-based design in high seismic risk is handled by specifying a reduced plastic hinge length. For low seismic risk, such as in SDC A or Seismic Zone 1, Type 2 connectors are permitted in plastic hinge regions without restriction because inelastic demand levels are expected to be minimal. Research is needed to investigate the behavior of mechanical splices with high strength steel reinforcement or larger bars projected into the capacity-protected element to offset the formation of plastic hinges away from the column-footing or column-pier cap interface. Extensive experimental research exists in the literature about the static, cyclic and impact load behavior of mechanical connectors tested in air (Einea, 1995; Jansson, 2008; Noureddine et al., 1996; Rowel et al., 2009; Haber et al., 2013; Tazarv and Saiidi, 2014). While the mechanical connectors were generally able to develop the tensile strength of the splice bars, reduced elongation/ductility capacity was consistently observed in relation to control rebars. It should also be mentioned that in some cases, fracture of the Type 2 mechanical connectors have been observed in the laboratory, even for connectors in which the reinforcement is not tapered or threaded. Effect of loading rate on the dynamic performance of headed bar connectors was found to be negligible in the tests conducted at the University of Nevada at Reno (Haber et al. 2013), but Rowell et al. (2009) reported that the performance of these connectors was adversely affected with high loading rates. Further research is needed to assess the dynamic response of headed bar connectors. Recent research by Tazarv and Saiidi (2016) has made possible to quantify the reduced ductility of mechanical splices and their implication in the behavior of plastic hinge regions. In addition because provisions for Type 2 connectors already exist in ACI and AASHTO design documents, this article receives a high rating for technology readiness. The following is a technology readiness form for this article:

NCHRP Project 12-102 81 Figure 5.5.3.8-1 Technology Readiness Evaluation for Article 3.6.4.3

NCHRP Project 12-102 82 5.5.3.9 Article 3.6.4.4 Type 2 Mechanical Connectors in Plastic Hinge Regions for SDCs C and D (Seismic Zones 3 and 4) There are many types of mechanical connectors commercially available, however, column-to-footing assemblies tested in the laboratory under cyclic loading mostly used headed rebar coupler (HC) and grouted sleeve coupler (GC) splices in the plastic hinge zones (Haber et al., 2013; Pantelides et al., 2014; Tazarav and Saiidi, 2014). Shear screw couplers have been found to exhibit a reduced strength and deformation capacity because of stress concentrations and premature bar fracture under the screws. Because of this and recognizing the strict limitations that AASHTO Guide Specifications for LRFD Seismic Bridge Design (2011) places on the use of mechanical couplers, HC and GC are the only mechanical connectors that are proposed as permissible for plastic hinge zones in SDCs C and D or Seismic Zones 3 and 4. This limitation is based on the assumption that a coupler that can develop the static ultimate strength of the bar can also provide good performance under cyclic loading. The assumption is justified by the fact that HC and GC have been shown to achieve both capabilities, while other coupler types have been able to provide neither. Research is needed to characterize the seismic performance of other types of mechanical connectors in or near plastic hinge region of columns. Column-to-footing connections tested in the laboratory were restricted to mechanical couplers of lengths smaller than 15 times the column longitudinal bar diameter. Because the design provisions included in these specifications are consistent with those test results, the same limitation, in addition to the strength requirement, is imposed on the length of mechanical connectors. Staggering of mechanical splices in plastic hinge regions is not permitted in the proposed guide specifications because at a given location the non-spliced longitudinal bars are not laterally supported by stirrups. Upon spalling of concrete cover under seismic induced load reversals, these unsupported bars are prone to buckling and subsequent premature fracturing as observed in the laboratory (Phillip and Hegemier, 2013). Staggering of splices may be possible if means of lateral support were to be developed, and their effectiveness proven, by testing, but that has not yet been done. Significant amount of research has been conducted on the dynamic response of mechanical connectors and on the cyclic quasi-static lateral load behavior of column-footings and column-cap beam connections incorporating connectors in the plastic hinge regions. Reductions in the score relate to the following reasons: 1) assemblies were tested under slow reversed cyclic loading so research is needed on the dynamic response of the connection under large strain reversals (shake table tests), 2) column-footing column-pier cap assembly tests were primarily based on two proprietary connectors (NMB by Splice Sleeve and HRC by Headed Reinforcement Corporation), 3) design recommendations are available but specifications have not been developed. It should be mentioned that the performance of the splice is also heavily dependent on proper installation of the mechanical connectors in the field. The following is a technology readiness form for this article:

NCHRP Project 12-102 83 Figure 5.5.3.9-1 Technology Readiness Evaluation for Article 3.6.4.4

NCHRP Project 12-102 84 5.5.3.10 Article 3.6.4.4.1 Forced-Based Design of Column Connections with Mechanical Connectors for Seismic Zones 3 and 4 Although mechanical connectors can be efficient for developing the ultimate tensile strength of spliced reinforcement, precast connections with these devices have been found to exhibit reduced ductility and energy dissipation capacity as compared to CIP connections with the same reinforcement and no mechanical connectors. This occurs because mechanical connectors are very stiff and thus compromise the column curvature capacity in the splice zone in addition to leading to rebar strain concentrations adjacent to the end of the coupler. Significant reduction in displacement ductility capacity has been observed even when the connectors are embedded into the capacity-protected element (Pantelides et al., 2014). Post failure assessment of column-to-footing assemblies tested in the laboratory (Haber et al., 2013 and Pantelides et al., 2014) showed that GC and HC in the plastic hinge region essentially remained elastic, while splicing reinforcement fractured. In all cases, however, displacement ductility capacity of the precast connection was smaller than the displacement ductility capacity of a reference CIP assembly. Analytical and experimental research has shown that the reduction is more significant for: 1) mechanical couplers placed in the column rather than in the pier cap or footing; 2) longer mechanical connectors, such as GC; 3) mechanical connectors placed in the column and adjacent to its interface with the pier cap or footing. Table 5.5.3.10-1 summarizes the measured displacement ductility capacity of several precast connections with mechanical connectors as compared to that of corresponding reference CIP connection tested in the laboratory. Table 5.5.3.10-1 Displacement Ductility Capacity for Several Precast Connections with Mechanical Couplers Connector ൬ࣆࡰ ࡼ/࡯ ࣆࡰ࡯ࡵࡼ ൰2 Ref.3 Type Length1 Embedding Element Distance from Interface GC 14.6db Footing 0 0.69 1 GC 14.6db Column 0 0.61 1 HC 3.6db Column 4.0 in. 0.88 2 1 Expressed in terms of column longitudinal bar diameters, db 2 Displacement ductility capacity of p/c connection to displacement ductility capacity of reference CIP. 3 Ref. 1: Pantelides et al., (2014); Ref. 2: Haber et al., (2013) The proposed provision of using a reduced response modification factor for connections with mechanical couplers is justified as follows. Consider the relations between elastic demand, inelastic demand, and response modification factor (R) in a force-based design approach (Figure 5.5.3.10-1). The displacement ductility demand of the nonlinear system is given by: ߤ஽ = ୼೘୼೤ (5.5.3.10-1) in which:

NCHRP Project 12-102 85 Δ௬ = ቀ୼೐ோ ቁΩ (5.5.3.10-2) where: Δ௠= peak displacement demand of the nonlinear system Δ௬= yielding displacement of the nonlinear system Δ௘= displacement demand of the linear elastic system R = response modification factor Ω =overstrength factor (= 1.3 in AASHTO LRFD Bridge Design Specifications (2014) F௬= yielding force of the nonlinear system F௘= force demand of the linear elastic system Substituting Eq. (2) into Eq. (1) leads to: ߤ஽ = ቀ୼೘୼೐ ቁ ோ ஐ (5.5.3.10-3) Separately applying this equation to a precast (PC) system with mechanical connectors and to a reference CIP system with the same reinforcement but no mechanical connector and taking the ratio between the two resulting expressions leads to: ൬ఓವು಴ఓವ಴಺ು൰ = ቀ౴೘౴೐ ቁ ು಴ ቀ౴೘౴೐ ቁ ಴಺ು ቀஐ ಴಺ು ஐು಴ ቁ ோು಴ ோ಴಺ು (5.5.3.10-4) Except for short period structures, AASHTO Guide Specifications for LRFD Seismic Bridge Design (2011) alludes to the equal displacement approach, such that the displacement demand of an elastic system is equal to the displacement demand of the corresponding inelastic system (Δm/Δe =1.0). For short period structures, on the other hand, it can be assumed that the ratio of (Δm/Δe) for the precast system is equal to that for the reference CIP system. With these considerations in mind, Eq. (4) leads to: ோು಴ ோ಴಺ು = ቀ ஐು಴ ஐ಴಺ುቁ ൬ ఓವು಴ ఓವ಴಺ು ൰ (5.5.3.10-5) Pantelides et al., (2014) and Haber et al., (2013) observed that the lateral load capacity of column- footing assemblies with mechanical connectors was slightly greater than the lateral load capacity of reference CIP assemblies. This is believed to be due to the shift in the plastic hinge zone associated to the stiff and strong connector elements. Because the ratio of overstrength factors between precast and cast-in- place systems may therefore be expected to be greater than one, Eq. (5) conservatively reduces to: ோು಴ ோ಴಺ು = ൬ ఓವು಴ ఓವ಴಺ು ൰ (5.5.3.10-6) For precast connections with long mechanical connectors, such as the GC, displacement ductility ratio has been measured to be as low as 60% of that for the reference CIP connection (Haber et al., 2013; Pantelides et al., 2014). For precast connections with short mechanical connectors, such as the HC, the

NCHRP Project 12-102 86 measured displacement ductility has been measured to be 88% of that for CIP (Haber et al. 2013) as shown in Table 5.5.3.10-1. Equation (3.6.4.4.1-1) in the proposed guide specifications corresponds to the selection of conservative values of: ఓವ ು಴ ఓವ಴಺ು (0.5 for CG and 0.8 for HC) to allow estimating the reduced R-factor of the precast connection with mechanical couplers in the plastic hinge region. Analytical research is needed to more strictly develop response modification factors, R, for Type 2 mechanical connectors in plastic hinge zones while recognizing the reduced displacement ductility capacity that has been observed in the laboratory. Experimental results for column-to-footing and column-to-pier cap assemblies tested by Pantelides et al., (2014) show that the seismic performance of the system improves when mechanical connectors are placed in the footing or pier cap rather than in the column. Although such configuration is encouraged in the proposed guide specifications, challenges associated with spatial conflict may arise. In that case, placing of the connector in the column is acceptable, provided that the spliced reinforcement is partially debonded into the pier cap or footing, because this has been found to prevent premature bar fracture, reduce spalling, and overall improvement to the seismic performance of the connection. Figure 5.5.3.10-1 Relation Between Elastic and Inelastic Displacement in a Force-Based Design Approach ܨ௘ ܨ௬ ܨ௘/ܴ ߂௬ ߂௘ ߂௠ ܴ ߗ

NCHRP Project 12-102 87 The following is a technology readiness form for this article: Figure 5.5.3.10-2 Technology Readiness Evaluation for Article 3.6.4.4.1

NCHRP Project 12-102 88 5.5.3.11 Article 3.6.4.4.2 Displacement-Based Design of Column Connections with Mechanical Connectors for SDCs C and D Provisions in this section are primarily based on the work by Tazarv and Saiidi (2014, 2015, and 2016), in which it is recognized that the seismic performance of columns with mechanical connectors in the plastic hinge region is inferior to that of conventional CIP. An approach was developed in that research to quantify the reduction in the displacement ductility capacity as summarized next. The displacement capacity of columns with mechanical connectors can be estimated using distributed plasticity (or fiber) models that account for the expected stress-strain relationships for concrete, reinforcement outside of the connector region, and reinforcement inside of the connector region. The latter can be obtained using the reinforcement stress-strain relationship but shifted to a reduced strain (ߝ௦௣) that accounts for the presence of the mechanical connector (Figure 5.5.3.11-1). The relation between ߝ௦௣ and the bar strain (ߝ௦) ensues by assuming that the connector is effectively rigid over a distance ߚܮ௦௣: ఌೞ೛ ఌೞ = ௅೎ೝିఉ௅ೞ೛ ௅೎ೝ (5.5.3.11-1) where: ܮ௦௣ = length of mechanical connector (in.); ܮ௖௥ = length of connector region = ܮ௦௣ + 2݀௕௟ ߚ = connector rigid length factor, as listed Table 6-1. The parameter ߚ was obtained from calibration of this numerical model with pullout test results involving mechanical connectors. Recommended values for the relevant mechanical connectors in the proposed guide specification are listed in Table 5.5.3.11-1. Table 5.5.3.11-1. Rigid Length Factor for Permitted Mechanical Connectors Mechanical Connector Type Rigid Length Factor, ߚ GC 0.65 HC 0.75

NCHRP Project 12-102 89 Source: Saiidi and Tazarav, 2016 Figure 5.5.3.11-1 Stress-strain Model for Mechanical Bar Splices Tazarav and Saiidi (2015) verified and calibrated this mechanically spliced column fiber model using available experimental results of column-footing assemblies. They then conducted a parametric study to investigate the effect of different factors on the displacement ductility capacity of columns. It was found that connector length, connector location, and type of connector significantly affect the displacement ductility capacity of mechanically spliced columns. Results from the analyses were synthetized into a best-fit equation for estimating the displacement ductility capacity of a precast column with connector, ߤ௉஼ , as a function of the displacement capacity of the conventional CIP column with the same reinforcement but with no connector, ߤ஼ூ௉: ఓು಴ ఓ಴಺ು = (1 − 0.18ߚ) ൬ ுೞ೛ ௅ೞ೛൰ ଴.ଵ଼ఉ (5.5.3.11-2) where: ܪ௦௣ = Because implementation of Eq. (2) requires calculating the displacement ductility capacity of a reference CIP system first, a more versatile approach was later suggested by Tazarv and Saiidi (2016) so that the plastic hinge model included in AASHTO Guide Specifications for LRFD Seismic Bridge Design (2014) could directly be used to estimate the displacement ductility of the precast system. As illustrated in Figure 5.5.3.11-2, this was accomplished by introducing a reduced plastic hinge length that accounts for the presence of the mechanical connector in the plastic hinge region: ܮ௣௦௣ = ܮ௣ − ൬1 − ுೞ೛௅೛ ൰ ߚܮ௦௣ ≤ ܮ௣ (5.5.3.11-3) distance from the top surface of footing or bottom surface of pier cap to the nearest end of the mechanical connector (in.). Bar region Coupler region Strain St re ss B ar re gi on B ar re gi on C ou pl er re gi on (a) Actual (b) Simplified L s p L c r βL sp (c) Rigid zone R ig id

NCHRP Project 12-102 90 where: ܮ௣௦௣= ܮ௣ = Figure 5.5.3.11-2 Actual and Idealized Curvature Diagram for Column with Mechanical Connector in Plastic Hinge Region. Figure 5.5.3.11-3 shows the reduced plastic hinge length factor (ܮ௣௦௣/ܮ௣) as a function of the location of the mechanical splice in the column (ܪ௦௣) for the relevant coupler device types and connector lengths (ܮ௦௣). No reduction occurs when the coupler device is placed outside of the analytical plastic hinge region. By contrast, the most significant reduction in plastic hinge length occurs for longer mechanical connectors (such as GC) placed at the end of the column and adjacent to the pier cap or footing (ܪ௦௣ = 0). Consistent with these, more significant reduction in displacement capacity has been observed in the laboratory for column-footing and column-pier cap assemblies with relatively long GC connectors at the column’s end. reduced analytical plastic hinge length to account for the presence of the mechanical connector (in.) analytical plastic hinge length for CIP members as defined in AASHTO Guide Specification for LRFD Seismic Bridge Design (2011) (in.)

NCHRP Project 12-102 91 Top: HC, Bottom: GC. Figure 5.5.3.11-3 Reduced Plastic Hinge Length to Account for Mechanical Splice in Plastic Hinge Region: 0.4 0.5 0.6 0.7 0.8 0.9 1.0 1.1 0.0 0.2 0.4 0.6 0.8 1.0 1.2 L ps p /L p Hsp/Lp HC (Lsp/Lp=0.15) HC (Lsp/Lp=0.3) 0.40 0.50 0.60 0.70 0.80 0.90 1.00 0 0.2 0.4 0.6 0.8 1 1.2 L ps p /L p Hsp/Lp GC (Lsp/Lp=0.3) GC (Lsp/Lp=0.5)

NCHRP Project 12-102 92 Table 5.5.3.11-2 shows comparisons of the measured displacement ductility capacity for three column- to-footing assemblies tested in the laboratory with the displacement ductility capacity calculated by Tazarav and Saiidi (2016). It is apparent that the reduced plastic hinge length approach provide adequate, although slightly unconservative, estimates of the displacement ductility capacity of the column-footing assemblies. Table 5.5.3.11-2. Comparison of Measured and Calculated Displacement Ductility of Columns with Mechanical Connectors using a Reduced Plastic Hinge Length Ref1 Connection Type Hsp (in.) Lsp (in.) M: Measured μD C: Calculated2 μD Ratio = C/M a) GC(β=0.65) 0 14.6 4.52 4.68 1.04 a) HC(β=0.75) 4 3.1 6.49 7.01 1.08 b) GC(β=0.65) 0 14.6 5.40 5.75 1.06 Source: 1: a): Haber et al., (2013), b): Pantelides et al., (2014) 2: using AASHTO Guide Specifications with a reduced plastic hinge length (Tazarv and Saiidi, 2016) While Eq. (5.5.3.11-3) was derived with the purpose of allowing to estimate the displacement capacity of columns with mechanical connectors, the expression could also be conservatively used when mechanical splices are placed in the capacity-protected element by taking Hsp = 0. Using this approach for a test specimen with mechanical connector in the footing (Pantelides et al., 2014) a displacement ductility capacity of μD = 5.75 is calculated (same as that listed in the third row of Table 55.5.3.11-2), while the displacement ductility capacity was measured in the laboratory was μD = 6.10. The following is a technology readiness form for this article:

NCHRP Project 12-102 93 Figure 5.5.3.11-4 Technology Readiness Evaluation for Article 3.6.4.4.2

NCHRP Project 12-102 94 5.5.3.12 Article 3.6.4.5 Debonding of Column Longitudinal Reinforcement for Mechanical Couplers in the Plastic Hinge Region The provisions in this article are intended to provide an improvement in the seismic behavior when bar connectors are placed at the interface of a precast column with a capacity-protected element. Debonding of column reinforcement into pier cap or footing has been associated with delay of rebar fracture and reduced spalling of adjacent capacity-protected elements (Belleri and Riva, 2012, Mashal et al., 2014, Pang et al., 2008). These behaviors lead to improved system ductility and energy dissipation capacity. For the GC column-to-footings connections tested by Pantelides et al., (2014), for example, debonding increased the measured displacement ductility capacity from 5.4 to 6.8. There is need for systematic research to assess the effect of debonding length, debonding materials, and location of debonding on the seismic performance of column-to-pier cap and column-to-footing assemblies. In order to estimate the minimum required debonding length, the model shown in Figure 5.5.3.12-1 was used as originally proposed by Pang et al. (2008). It is assumed that all connection rotation, φ, concentrates at the interface with the footing so that the column behaves as a rigid body. Under that premise, the maximum strain in the column longitudinal reinforcement, ߝ௠௔௫, is given by: ߝ௠௔௫ = ஦(ௗି௖)௅೏೐್ (5.5.3.12-1) where: c = neutral axis depth at the column-footing interface d = distance from the compression end of the column to the tension-most layer of reinforcement ܮௗ௘௕ = debonding length of column longitudinal reinforcement into footing Figure 5.5.3.12-1 Model to Determine the Required Debonded Length in a Column-to-Footing Connection Dc Ldeb d c lac

NCHRP Project 12-102 95 Limiting the tensile strain in the steel to ߝ௠௔௫ = 0.04 for a drift ratio of three percent (φ= 0.03 rad), the debonding length is calculated to be: ܮௗ௘௕ = ଴.଴ଷ(ௗି௖)଴.଴ସ (5.5.3.12-2) Assuming a neutral axis depth c = d/3 and taking d = 0.8Dc leads to the first part of Eq. 3.6.4.5-1 in the proposed guide specifications, which is given by: ܮௗ௘௕ ≥ max ൜ 0.4ܦ௖5000ߝ௬݀௕௟ (5.5.3.12-3) As seen in the second part of Eq. (5.5.3.12-3), the debonding length is not to be taken less than the “strain penetration” component of the theoretical plastic hinge region for CIP in AASHTO Guide Specifications for LRFD Seismic Bridge Design (2011) but written in terms of the yielding strain (so that the equation is dimensionally compatible). Because the average (vertical) joint shear stress depends on the length of column reinforcement embedded into the pier cap or footing, research is needed to evaluate the effect debonding of column longitudinal reinforcement on joint behavior. Until this knowledge gap is addressed, the proposed approach is to neglect the debonded length in joint shear calculations. Although extensive research is available, no specifications were found that deal with debonding of column reinforcement into beams or footings. Because implementation of the specific technology for bridge connections was not found either and systematic research on the effect of debonding, the article scores slightly below the threshold value for inclusion in the specifications. However, it is recommended that this provision be kept as a simple alternative for improving the seismic performance of connections The following is a technology readiness form for this article:

NCHRP Project 12-102 96 Figure 5.5.3.12-2 Technology Readiness Evaluation for Article 3.6.4.5

NCHRP Project 12-102 97 5.5.3.13 Column Connections with Mechanical Connectors Not Included in the Proposed ABC Guide Specifications Because of Low Technology Readiness Score This section presents examples of connections that were evaluated in this project but that were not included in the proposed guide specifications because of the low technology readiness score. Over the last decade, experimental and numerical research has been conducted at the University of Nevada Reno with the purpose of developing column connections with mechanical couplers that could reasonably emulate the behavior of reference cast-in-place assemblies. Some alternative details that were found to provide adequate seismic response in terms of strength, deformation capacity, and energy dissipation capacity are described next: Column Connection with Up-set HC and Transition Bar The HC connection incorporates an open region used to provide working space for the installation of transition bars (Figure 5.5.3.13-1). Installation of these bars, originally intended to work a replaceable fuses following earthquake actions, requires tight tolerance, more construction time, and field adjustments. Based on experimental results and numerical modeling, Haber et al. (2013) found that the design of the connection does not require any special analysis compared to a CIP connection to account for the presence of the coupling devices. These splices have little effect on the force-displacement behavior of the column and thus render emulative behavior of CIP in terms of displacement and energy dissipation capacity as shown in Figure 5.5.3.13-2. Source: Haber et al. 2013 Figure 5.5.3.13-1 Emulative HC Connection with Transition Bar

NCHRP Project 12-102 98 Source: Haber et al. 2013 Figure 5.5.3.13-2 Comparison of HC Emulative Column-Footing Assembly Relative to Reference CIP Column Connection with Grouted Sleeve Coupler with Debonded Pedestal (GCDP) Experiments with half-scale column-to-footing assemblies showed that bridge columns with GCs shifted away from the column’s end can reasonably emulate the seismic response of CIP counterparts (Tazarv and Saiidi, 2014) in terms of energy dissipation and displacement capacity. Consequently, these column types can be designed similarly to conventional columns according to existing design codes for CIP. As shown in Figure 5.5.3.13-3, the shifted splice location is achieved through the use of a CIP pedestal through which coupling bars are debonded to prevent strain concentrations. Source: Tazarv and Saiidi 2014 Figure 5.5.3.13-3 Details of a GCDP Splice

NCHRP Project 12-102 99 Source: Tazarv and Saiidi 2014 Figure 5.5.3.13-4 Comparison of GCP Emulative Column-Footing Assembly Relative to Reference CIP Haber et al (2013) developed an analytical plastic hinge model, which was calibrated with test results, and conducted a parametric study to determine the effect of splice length and splice location on the seismic response of column-footing assemblies. It was found that pre- and post-yield stiffness of mechanical splices should be considered in ductility calculations when the splice length to bar diameter ratio was greater than 4.0. It was also found that placing the splices at half the column diameter from the column-footing interface was conducive to a reasonably emulative seismic behavior as shown in Figure 5.5.3.13-4. Because the details are very specific, implementation in the field was not found in the literature. In addition, the need to shore/grout/and form in order to cast the grout around transition bars in the HC connection and the use of a CIP pedestal for the GCP connection pose the question on whether these connection details offer any meaningful time saving in construction relative to CIP. Finally, durability can be a concern for both connections when they are located below grade. As a result of these, the item receives a low technology readiness factor and it is not recommended in the proposed ABC Guide Specifications. The following is a technology readiness form for this article:

NCHRP Project 12-102 100 Figure 5.5.3.13-5 Technology Readiness Evaluation for Column Connections with Mechanical Connectors

NCHRP Project 12-102 101 5.5.3.14 Article 3.6.5 Grouted Duct Connections A grouted duct connection shares some characteristics with a grouted sleeve mechanical coupler but differs in its intended function. In both cases a bar is secured by grout into a surrounding metal sleeve. However, a grouted sleeve connects two collinear bars in a butt-splice, so the sleeve itself needs to have at a tension strength higher than that of the connected bars. By contrast, a grouted duct connection anchors a single bar into a body of concrete by bond stresses, so the duct has no tension strength requirement. The tension force in the bar may be transferred into another bar that is lap spliced to the duct, but those requirements are separate Generally, provisions in subsections of Article 3.6.5 are based on design specifications and supporting research included in FHWA Report No. HIF-13037 (Marsh et al., 2013) and in NCHRP Report No. 681 (Restrepo et al., 2011). Precast test specimens with grouted duct connections that were detailed using current design standards performed well under cyclic loading and were considered to be emulative of CIP counterparts for which plastic hinging occurs at the column ends while joint zones remain essentially uncracked (Belleri and Riva, 2012; Mashal et al., 2014; Pang et al., 2010). As an example of this, Figure 5.5.3.14-1 shows the lateral force-versus-lateral displacement response of column-pier-cap assembly with grouted ducts reported by Restrepo et al., (2011). Similar to a reference CIP column, the precast specimen exhibited a stable hysteretic behavior without significant strength degradation up to displacement ductility of 8. Post-test inspection of the specimen revealed that the core and bedding layer remained primarily intact with several column bars buckling and two bars fracturing at ultimate displacement. Because grouted ducts have been investigated in the form of both elements and assemblies, in addition to being successfully implemented in several bridge projects, this general article scores high in technology readiness. The following is a technology readiness form for this article: Source: Restrepo et al., 2011 Figure 5.5.3.14-1 Cyclic Lateral Load Response of Column-Pier Cap Assembly with Grouted Ducts

NCHRP Project 12-102 102 Figure 5.5.3.14-2 Technology Readiness Evaluation for Article 3.6.5

NCHRP Project 12-102 103 5.5.3.15 Article 3.6.5.1 Minimum Development Length of Reinforcing Steel Results from pullout tests have demonstrated that corrugated steel ducts serve to arrest splitting cracks and increase local confinement and shear transfer from the bar to surrounding concrete. This effectively translates into higher concrete-steel bond strength and a reduction in the anchorage length required to develop the tensile strength of the reinforcement inside the ducts. The development length equation included in the proposed guide specification correspond to that in FHWA Report No. HIF-13037 (Marsh et al., 2013). The expression is based on the results from monotonic pullout tests conducted by Eberhard et al., (2009) and is based on the assumption that the tensile strength of the bar is 1.4 times the expected yield strength of the material to account for dynamic effects. The development equation suggested in NCHRP Report No. 681 (Restrepo et al., 2011) was not adopted in the proposed provisions because: 1) supporting pullout tests (Matsumoto et al., 2008) mostly correspond to unconfined joints of relevance to TxDOT construction with non-seismic detailing, 2) the development length is given as proportional to the inverse of the grout’s compressive strength, f’cg, as opposed to its tensile strength (square root of f’cg) which is the traditional form in design specifications/codes and conventional knowledge about bond. As shown in Figure 5.5.3.15-1, for an expected yield stress of ௬݂௘ = 68 ݇ݏ݅, development length calculated with the equation in NCHRP Report No. 681 (Restrepo et al., 2011) is conservative as compared to what is obtained with the expression for “Minimum Development Length of Reinforcing Steel for SDCs C and D” in AASHTO Guide Specifications for LRFD Seismic Bridge Design (2011), especially for low and high grout compressive strengths. It is apparent, therefore, that no recognition is given in the former to the increase in bond strength due to the corrugated ducts. By contrast, the proposed development length equation amounts to a 18% reduction in the development length as compared to what can be calculated using with AASHTO. As illustrated in Figure 5.5.3.15-2, tensile forces at the connection are transferred to the duct and surrounding concrete through bond. In order to prevent slippage of the duct itself it is required that the bond strength of the duct embedded into concrete should be at least one third of that for deformed bars because a minimum duct to bar diameter ratio of three is required in the proposed guide specifications. Until research on the bond characteristics of ducts embedded in concrete becomes available, the conservative approach of requiring corrugated steel pipes confirming to ASTM A760 is adopted. The latter are known to have better bond characteristics compared to ducts used in post-tensioning. For low seismic risk, the required amount of spiral in joint regions may be less than that used in test specimens leading to the development length equation in this article (Equation 3.6.5.1-1). A conservative approach is adopted in the proposed guide specifications to compensate for this by requiring that the anchorage length also satisfy the development length requirements in AASHTO LRFD Bridge Design Specification for SDCs A and B or for Seismic Zones 1 and 2. Research is needed to quantify the effect of joint confinement on the bond behavior of grouted ducts.

NCHRP Project 12-102 104 0 5 10 15 20 25 30 35 40 4 5 6 7 8 9 10 11 12 13 14 15 l ac /d bl f'cg (ksi) NCHRP 681 (2011) AASHTO Guide Specs (2011) Proposed Figure 5.5.3.15-1 Comparison of Development Length Equations for fye= 68 ksi Figure 5.5.3.15-2 Tensile Load Transfer Mechanism at Connection Multiple pullout tests, including static and cyclic, have been conducted in the last fifteen years. Design specification and equations are available in reports FHWA-HIF-13037 and NCHRP 681. This type of connection has been implemented in multiple bridges not only in non-seismic regions (Texas and Florida) but also in seismic regions (California and Washington). Because of all these reasons, this item receives a high score for technology readiness. The following is a technology readiness form for this article: Bedding layer T Bar-to-grout bond Duct-to- concrete bond Grout

NCHRP Project 12-102 105 Figure 5.5.3.15-3 Technology Readiness Evaluation for Article 3.6.5.1

NCHRP Project 12-102 106 5.5.3.16 Article 3.6.5.2 Splicing of Longitudinal Reinforcement The requirements proposed in this section already exist in AASHTO LRFD Bridge Design Specifications (2014) as “Splices of Bar Reinforcement”. Tension force transfer should occur from the embedded reinforcement to the duct and from the duct to the splicing reinforcement outside. Column-to- pier cap grouted duct connections with spliced longitudinal reinforcement were tested by Pang et al. (2010). The grout inside the duct benefits from the confinement effect of the duct, but the lap splice between the outside of the duct and any bar next to it does not, so the development lengths may differ. For the “outside” bar, the column spiral may affect the development length, because it is doing the confining of that splice. In the test specimens, precast columns had large projecting bars that fit into the ducts of a precast pier cap. The protruding large bars were spliced with the column (smaller) longitudinal reinforcement. The lateral force-displacement response and damage progression in the precast connection was found to be similar to that of a reference cast-in-place concrete connection. For connections in which large bars that are spliced with the column reinforcement and extended into the pier cap of footing, there is need to investigate the effect of the additional and non-developed (column) bars in the plastic hinge zone. As pointed out by Pang et al., (2008), strain concentrations can be expected because of the additional longitudinal bars and this could influence the ductility capacity of the column-to-footing or column-to-pier cap assembly. The following is a technology readiness form for this article:

NCHRP Project 12-102 107 Figure 5.5.3.16-1 Technology Readiness Evaluation for Article 3.6.5.2

NCHRP Project 12-102 108 5.5.3.17 Article 3.6.5.3 Debonding of Column Longitudinal Reinforcement This article is intended to provide the designer with an optional detail for improved seismic behavior of the column-to-pier cap or column-to-footing connection with grouted ducts. Laboratory tests have on column-to-cap and column-to-footing connections have shown that debonding may prevent premature bar fracture, reduce spalling of capacity-protected elements, and delay cyclic strength degradation (Mashal et al. 2014, Belleri and Riva 2012, Pantelides et al., 2014). However, no study was found in the literature focused on systematically assessing the effects of debonding on the behavior of connections and/or the development of specifications to achieve improved seismic performance. Although much research exists, no specifications were found in the literature that deal with debonding of column reinforcement into pier cap or footings. Because documented implementation of the specific technology was not found either, the article scores slightly below the threshold value for inclusion in the proposed guide specifications. However, the provision is only presented as optional and therefore inclusion in the proposed design specifications may be justified as a result. The following is a technology readiness form for this article:

NCHRP Project 12-102 109 Figure 5.5.3.17-1 Technology Readiness Evaluation for Article 3.6.5.3

NCHRP Project 12-102 110 5.5.3.18 Article 3.6.5.4 Bedding Layer Multiple precast connections reported in the literature incorporated the use of a bedding layer to facilitate construction of the test specimens (Belleri and Riva 2012; Mashal et al., 2014; Restrepo et al., 2011). Provisions about maximum thickness and reinforcement of bedding layer are identical to those in the design specifications provided in NCHRP Report 681 (Restrepo et al., 2011). The use of fiber reinforcement in the bedding layer was not explicitly mentioned in the proposed ABC Guide Specifications but the following discussion is presented for information purposes: Fiber reinforcement has been used to improve the properties of grout in several projects. No study is known of that has provided a numerical evaluation of its advantages, but anecdotal references to its advantages are common. A bedding layer under a column is typically thin compared to its lateral dimensions, and that geometry provides most of the layer with significant lateral confinement and hence compressive strength. Under concentric compression, the load capacity of the layer is usually sufficient, even without fibers. The confinement offered by the fibers is likely to be less effective than that provided by the friction on the upper and lower surfaces of the bedding layer, on the basis of stiffness if nothing else, so the fibers confer no noticeable advantage. However, the edge of the layer is more vulnerable, because it does not enjoy the same level of confinement from friction as does the interior. If the column rocks on the bedding layer, its edge experiences high local stress and starts to break up. Under those circumstances, fibers provide improved performance, because they hold the grout together and inhibit its escape from the bedding layer. Consensus is lacking on the choice of design strength for the bedding layer material. Some authorities advocate that the nominal grout strength be less than that of the adjoining concrete elements, but that it should be fiber-reinforced to provide toughness, in order to capacity-protect the concrete, which is more difficult to repair. Even if the nominal strength of the grout material is less than that of the concrete, the grout layer will not crush under the column for the reasons discussed above. Others argue that any connection, including a bedding layer, should, on principle, always have a strength higher than that of the adjoining members. Fibers tend to impede the flow of fresh grout in constricted spaces, so fiber-based materials may be better installed by “buttering” down a layer prior to placing the column than by pouring or injecting it after column installation. For example Steuck et al. (2009) found that the grout that they used to grout bars in ducts resulted in lower bond strengths when it was fiber-reinforced. Part of the shortfall was attributed to difficulties in compaction. Several fiber-based materials may be considered for a bedding layer. Fibers may be added to packaged grouts, but dosages should be established in trial mixes prior to use. Engineering Cementitious Composites, or ECC, (Saiidi et al. 2009, Billington and Yoon, 2004) have been successfully used in the body of the column to improve toughness, and could also be considered for the bedding layer. In HYFRC, the concrete gains tensile strength and ductility from a strategic combination of different fiber types and sizes (Blunt and Ostertag, 2009). The potential uses are similar to those of ECC. Both materials use polymer fibers. UHPC (Graybeal, 2014) has high strengths of all types, but the tension and bond strengths are provided largely by the heavy fiber loading. Those fibers are typically steel, small diameter, and very high strength (300 – 400 ksi ultimate strength).

NCHRP Project 12-102 111 Because bedding layers have commonly used in the assembly of precast elements in the field and in experimental research, a high technology readiness score is assigned to this article. The following is a technology readiness form for this article: Figure 5.5.3.18-1 Technology Readiness Evaluation for Article 3.6.5.4

NCHRP Project 12-102 112 5.5.3.19 Article 3.6.5.5 Development of Deformed Steel Bars in Corrugated Steel Ducts Using UHPC Provisions in Article 3.6.5.5 are based on design specifications and supporting research by Tazarv and Saiidi (2013). Static and seismic testing demonstrated the emulative nature of these connections. Tests were completed on field-cast UHPC with 2-percent (by volume) steel fiber-reinforced, compressive strengths above 19 ksi and bar sizes ranging from No. 8 to No. 11. The research demonstrated that the bars could be developed within the duct. This connection is also similar to the provisions under Article 3.6.5.1. The only difference being the material used to develop the bar. For this reason, a technology readiness evaluation was not completed for this article.

NCHRP Project 12-102 113 5.5.3.20 Article 3.6.6 Pocket Connections Provisions in subsections of Article 3.6.6 are largely based on design specifications and supporting research included in NCHRP Report No. 681 (Restrepo et al., 2011) and in Tazarv and Saiidi (2015). A pocket is a prefabricated recess in CIP or precast element that is designed to receive reinforcing from a column and then be filled with CIP concrete or grout. Precast test specimens with pocket connections detailed using current design standards performed well under cyclic loading and were considered to be emulative of CIP counterparts for which plastic hinging occurs at the column ends while joint zones remain essentially uncracked. This is illustrated in Figure 5.5.3.20-1 which shows the lateral force-versus- lateral displacement response of a column-pier cap assembly with pocket connection reported by Restrepo et al., (2011). Similar to a reference CIP, this precast specimen exhibited a stable hysteretic behavior without significant strength degradation for displacement ductility of up to 8. Post-test inspection of the specimen revealed that two bars had fractured after buckling. Test specimens corresponded to interior joints in multi-column systems so further research is needed on the behavior of exterior (knee) joints with pocket connections. Source: Restrepo, et al. 2011 Figure 5.5.3.20-1 Cyclic Lateral Load Response of Column-Pier Cap Assembly with Pocket Connection Examples of implementation of pocket connection in column-to-pier cap connections for non-seismic regions include the Squaw Creek Bridge in Iowa, and the Escambia Bay Bridge in Florida (Wright, 2010). In addition three bridges at the Abu Dhabi’s Khalifa Port Industrial Zone (low seismic risk zone) used a CIP plug into the pocket of a pier cap (Karapiperis et al. 2010). Pockets formed with corrugated steel pipes were successfully used in the abutments of Boone County Bridge in Iowa (Wipf et al., 2009) and Madison County Bridge in Iowa (Phares et al., 2009) but these did not incorporate any seismic detailing.

NCHRP Project 12-102 114 As required with any of the connections listed in the proposed ABC Guide Specifications, the designer is responsible to identify the load path and provide proper detailing for force transfer from adjacent elements in accordance to AASHTO design provisions. For a pocket connections under an extreme condition in which the joint is cracked, for example, the load transfer mechanism may be as represented in Figure 5.5.3.20-2a with a simplified strut-and-tie model. In that case, the node on the upper left (identified with a broken circle) can become a weak point if joint detailing provisions were not used to ensure that a strut on the left of the node. Similarly, if the bottom reinforcement of the pier cap had been moved to either side of the prefabricated pocket then properly detailing is required so that struts can form at the node on the upper right of Fig. 5.5.3.20-2b. The latter was not an issue in the tests conducted by Restrepo et al. (2011) because pier cap longitudinal reinforcement crossed the prefabricated pocket directly over and under the corrugated steel pipe. However, the configuration with longitudinal bars on either side of the pocket may be more attractive for construction. The provisions “Joint Design for SDCs C and D” included in the AASHTO Guide Specification for LRFD Seismic Bridge Design (2011) are intended to cover these situations of providing adequate force transfer mechanisms. Figure 5.5.3.20-2 Strut-and-Tie Model for Load Transfer Across Cracked Joint The following is a technology readiness form for this article:

NCHRP Project 12-102 115 Figure 5.5.3.20-3 Technology Readiness Evaluation for Article 3.6.6.1

NCHRP Project 12-102 116 5.5.3.21 Article 3.6.6.2 Formed Pocket and Fill The seismic performance of pocket connections for column-to-pier cap assemblies has only been assessed in laboratory tests involving the use lock-seam corrugated steel pipes to form the pockets (Restrepo et al., 2011; Mehrsoroush and Saiidi, 2014 & 2016, Wipf et al., 2009). Proposed guide specifications are consistent with the details included in those experimental programs. No other alternatives are given to form the pockets because: 1) lack of experimental and analytical research, 2) corrugated steel pipes are readily available commercially at competitive cost, 3) in addition to serving as forms for the pockets corrugated steel pipes provide confinement and improve joint integrity. Corrugated pipes with lock-seam are specifically required in the proposed provisions also for consistency with laboratory specimens leading to these guide specifications. Corrugated pipes with welded seams might prove satisfactory, but they are not permitted until research can demonstrate that these can provide a reliable bar load transfer without compromising ductility. Experimental results included in NCHRP Report 681 (Restrepo et al., 2011) suggest that a properly sized steel pipe may eliminate the need for transverse reinforcement in the joint region. Tazarv and Saiidi (2015) also suggested providing additional spiral reinforcement along half depth of the pocket adjacent to the column-to-pier cap interface as a measure to enhance joint integrity. In the tests leading to that recommendation, however, it was not reported whether corrugated steel pipes in the connection had been sized to provide joint confinement, so the need for spiral reinforcement around the pocket remains unclear. The use of supplementary hoop near the top and another hoop near the bottom of the corrugated steel pipe was proposed in NCHRP 681 as a measure to prevent unraveling of the pipe itself. But tests by Mehrsoroush and Saiidi (2014) showed that portions of a spiral around the pocket and away from the column-footing or column-pier cap interface experience insignificant strains during the applied load cycles. As a result, in the proposed guidelines it is optional to provide confinement spiral around pockets that are formed with properly sized corrugated steel piles. The following is a technology readiness form for this article:

NCHRP Project 12-102 117 Figure 5.5.3.21-1 Technology Readiness Evaluation for Article 3.6.6.2

NCHRP Project 12-102 118 5.5.3.22 Article 3.6.6.3 Minimum Development Length of Reinforcing Steel for SDCs C and D (Seismic Zones 3 and 4) The development equation in NCHRP Report No. 681 (Restrepo et al., 2011) is based on the results of only four pullout tests with straight reinforcing bars into single and double-line grouted pockets. The test specimens were intended to be representative of connection details used in TxDOT so no corrugated steel pipe was used to form the pocket and joints did not incorporate seismic detailing. These test results were first reported in Matsumoto et al., (2001) and then in Matsumoto et al. (2008). Different development length equations have been suggested in each of those publications as follows: • Matsumoto et al., (2001): ݈௔௖ ≥ ଴.଻ௗ್೗௙೤೐ට௙೎೒ᇲ (5.5.3.22-1) • Matsumoto et al., (2008): ݈௔௖ ≥ ଷௗ್೗௙೤௙೎೒ᇲ with ௖݂௚ ᇱ limited to 6.5 ksi or less, bars #11 or smaller (5.5.3.22-2) • NCHRP Report 681 (Restrepo et al., 2011): ݈௔௖ ≥ ଶ.ଷௗ್೗௙೤೐௙೎ᇲ with ௖݂ ᇱ limited to 7.0 ksi or less, bars #11 or smaller (5.5.3.22-3) Figure 5.5.3.22-1 shows a comparison of the results obtained with these three equations for assumed reinforcement yielding stress of fy = fye = 68 ksi. The development length equation from AASHTO Guide Specifications for LRFD Seismic Bridge Design (2011) is included as a reference together with the results obtained with the expression included in the proposed guide specification. Explanation of the differences in the results is presented next. Figure 5.5.3.22-1 Comparison of Development Length Equations

NCHRP Project 12-102 119 Equation (1) was not intended for seismic applications and so it produces a shorter development length compared to the other two expressions. The discrepancy between Equation (5.5.3.22-2) and Equation (5.5.3.22-3) is explained in NCHRP 681 report (Restrepo et al., 2011) as follows: “(previous) anchorage equations were modified by removing a 0.75 factor that accounted for extensive splitting cracks at reentrant corners of grout pockets. Such cracking did not develop for the cylindrical-shaped cap pocket connections for CPFD (cap pocket full ductility specimen) and CPLD (cap pocket limited ductility specimen) that used steel pipes as stay-in-place forms”. The 0.75 factor must refer to the bond strength so that “removing the factor” from the denominator (proportional to bond strength) in Eq. (2) means multiplying the nominator by 0.75 (thus giving the 2.3 coefficient in Eq. 5.5.3.22-3). The NCHRP Report 681 development length equation for pocket connections (Eq. (5.5.3.22-3) above) is not adopted in the proposed specifications because that expression is given as proportional to the inverse of the concrete compressive strength, ௖݂ᇱ, as opposed to its tensile strength (ඥ ௖݂ᇱ) which is the traditional form in design specifications/codes and conventional knowledge of concrete-steel bond behavior. It is believed that proposing an equation with an unconventional form can only be warranted if more test data is available to support the change. The expression included in the proposed specification can be further justified by the following arguments: 1. Calculated development length is 40% larger than what is obtained with the original expression proposed by Matsumoto et al (2001). A 1.4 seismic modification factor for development length is conservative and consistent with what codes have traditionally prescribed. 2. Calculated development length is within 15% of what is obtained from using the expression proposed in NCHRP 681 report for a wide range of values of concrete compressive strength ௖݂ᇱ (from 4 to 12 ksi). 3. Calculated development length is 50% larger than what is obtained with the expression for grouted ducts in the proposed guide specifications (Article 3.6.5.1). It should be mentioned that the measured average bond strength normalized by ට ௖݂௚ᇱ was about the same for the grouted duct and the pocket connections reported by Matsumoto et al. (2008). Thus, the 1.5 factor represents a reasonable margin of safety given the limited amount of data on pullout capacity of bars embedded in pocket connections. Clearly, there is need for research to investigate the monotonic, cyclic, and group pullout tests of bars embedded in pockets formed with corrugated steel pipes. The proposed development length equation in the design specifications is expected to be conservative. For low seismic risk where the required amount of spiral in joint regions may be small a conservative approach is adopted in the proposed guide specifications to compensate for this by requiring that the anchorage length also satisfy the development length requirements in AASHTO LRFD Bridge Design Specification for SDCs A and B or for Seismic Zones 1 and 2. The following is a technology readiness form for this article. Because of the limited amount of research on the bond strength of reinforcement embedded in pockets (and the absence of results for pockets formed with corrugated ducts) this item scores slightly below the acceptance criterion. The proposed development length equation is believed to be conservative and could then make this provision acceptable until more research becomes available. This is also based on interest in industry for this type of connection.

NCHRP Project 12-102 120 Figure 5.5.3.22-2 Technology Readiness Evaluation for Article 3.6.6.3

NCHRP Project 12-102 121 5.5.3.23 Article 3.6.6.4 Corrugated Steel Pipe Thickness Requirements in this article are based on the recommendations given in NCHRP Report No. 681 (Restrepo et al., 2011). The intent of the provision is to ensure that the forming corrugated steel pipe could provide at least the same level of confinement as that provided by the minimum joint shear (transverse) reinforcement for SDCs C and D in a CIP connection. Although minimum joint transverse reinforcement ratio is not required for other design categories or zones, the same equations are conservatively proposed to allow estimating the corrugated steel pipe thickness in those cases. In order to derive an expression for the required pipe thickness consider a concrete column or joint segment as shown in Figure 5.5.3.23-1. Source: Restrepo et al., 2011 Figure 5.5.3.23-1 Confinement Provided by Spiral and Corrugated Steel Pipe Alternative forms of transverse reinforcement for the concrete element consist of a steel spiral or a corrugated steel pipe both with an average diameter ܦ௖௣ᇱ . In the first case, the maximum design confinement force per unit of length, ܨு, that spiral transverse reinforcement can provide is given by: ܨு = ஺ೞ೛௙೤೓௦ = ఘೞ஽೎೛೑೤೓ ᇲ ସ (5.5.3.23-1) where: Asp = spiral of cross-sectional area ௬݂௛ = nominal yield stress of transverse reinforcement in emulated CIP connection (ksi). s = spiral pitch ߩ௦ = required spiral volumetric ratio within the pier cap In the second case (Figure 5.5.3.23-1b) the maximum design confinement force per unit of length, ܨ௣௛, of the helical corrugated steel pipe may be estimated as: ܨ௣௛ = ௬݂௣௜௣௘ݐ௣௜௣௘cos (ߙ) (5.5.3.23-2) where: ௬݂௣௜௣௘ = nominal yield stress of corrugated steel pipe (ksi)

NCHRP Project 12-102 122 ݐ௣௜௣௘ = thickness of corrugated steel pipe (in.) ߙ = angle between horizontal axis of receiving member and pipe helical corrugation or lock seam (deg) Equating the right sides of Equations (1) and (2) and rearranging gives the thickness of the pipe as a function of the required spiral volumetric ratio within the pier cap: ݐ௣௜௣௘ = ఘೞ஽೎೛ ᇲ ௙೤೓ ସ௙೤೛೔೛೐ୡ୭ୱ (ఈ) (5.5.3.23-3) Eq. (3) is a rearranged version of an expression first derived by Restrepo et al. (2011). It renders the required corrugated steel pipe thickness as a function of the yielding strength of the spiral reinforcement that the pipe itself is substituting. As implied in that report, it is proposed that ௬݂௛ = 60 ݇ݏ݅ be used in the calculation as this corresponds to most commonly used (grade 60) transverse reinforcement. For SDCs C and D, AASHTO Guide Specifications for LRFD Seismic Bridge Design (2011) specifies the minimum value of ߩ௦ as a function of the calculated principal tension stress in the joint, pt. When ݌௧ < 0.11ඥ ௖݂௣ᇱ the joint is considered as uncracked and the minimum spiral volumetric ratio is specified as ߩ௦ = ଴.ଵଵට௙೎೛ᇲ ௙೤೓ . When ݌௧ > 0.11ඥ ௖݂௣ ᇱ the joint is considered as cracked and the minimum required spiral volumetric ratio is ߩ௦ = ଴.ସ஺ೞ೟ ௟ೌ೎మ . For low seismic design categories and zones, in which principal stresses in the joint are not required according to AASHTO provisions, it is proposed that the minimum ߩ௦ that is required for uncracked joints in SDCs C and D be used to conservatively estimate the thickness of corrugated pipe using Equation (5.5.3.23-3). Substitution of ߩ௦ = ଴.ଵଵට௙೎೛ᇲ ௙೤೓ into Eq. (3) above leads to equation in the proposed guide specifications. For forced-based design in Seismic Zones 3 and 4 where joint principal tensile stress is not calculated, it is proposed that the thickness of the pipe be calculated using Equation (3) with ߩ௦ taken as the maximum between the minimum required for uncracked joints (ߩ௦ = ଴.ଵଵට௙೎೛ᇲ ௙೤೓ ) and the minimum required for cracked joints in SDCs C and D (ߩ௦ = ଴.ସ஺ೞ೟ ௟ೌ೎మ ). This provision is based on observations made by Restrepo et al. (2011) that in some cases the AASHTO Guide Specifications for LRFD Seismic Bridge Design (2011) minim required ߩ௦ for uncracked joints can exceed the minimum required ߩ௦ for cracked joints The following is a technology readiness form for this article. Although corrugated steel pipes were used in column-to-pier cap pocket connections tested in the laboratory (Restrepo et al., 2011), thickness of the pipe was not a test variable. There is need to investigate the seismic behavior of pocket connections with different types and configurations of the pipe used to form the pocket. There is also need to specifically investigate the conditions under which corrugated steel pipe can serve as a substitute for joint confinement reinforcement. Overall score for this article is marginally higher than the minimum required for inclusion in the guide specifications.

NCHRP Project 12-102 123 Figure 5.5.3.23-2 Technology Readiness Evaluation for Article 3.6.6.4

NCHRP Project 12-102 124 5.5.3.24 Article 3.6.6.5 Bedding Layer Multiple precast connections reported in the literature incorporated the use of a bedding layer to facilitate construction of the test specimens (Belleri and Riva 2012; Mashal et al., 2014; Restrepo et al., 2011). Provisions about maximum thickness and reinforcement of bedding layer are identical to those in the design specifications provided in NCHRP Report 681 (Restrepo et al., 2011). The following is a technology readiness form for this article. Because bedding layers have commonly used in the assembly of precast elements in the field and in experimental research, a high technology readiness score is assigned to this article.

NCHRP Project 12-102 125 Figure 5.5.3.24-1 Technology Readiness Evaluation for Article 3.6.6.5

NCHRP Project 12-102 126 5.5.3.25 Article 3.6.6 Abutment-to-Pile Pocket Connections The provisions in this article are based on experimental research with scaled precast integral abutments conducted by Wipf et al., (2009). Test results showed that the presence of a CMP filled with concrete had no effect on the behavior of the abutment-to-pile connection and that neglecting the contribution of the corrugated steel pipe to the tension capacity of the connection is conservative. It is recommended therefore that detailing of the connection follow that required for conventional CIP abutments according to AASHTO provisions. Based on engineering judgement it is proposed that the pullout capacity of the corrugated pipe be calculated using its contact (shaft) area and the unit two-way shear strength of concrete. For a 24-inch pile and an approximate unit shear capacity of 0.13ඥ ௖݂ᇱ = 0.25 ݇ݏ݅, a reduced pullout capacity of 0.75 × ሾ0.25݇ݏ݅ × (ߨ × 24"×12")ሿ = 170 ݇݅݌ݏ per foot of embedment is obtained. Further research is needed to verify these assumptions. Depending on how the construction technique, the connection can be classified as a pocket or socket. In both cases a prefabricated recess is introduced in the precast or cast-in-place abutment. In a pocket connection the recess receives the projecting reinforcement of a pile, while in the socket connection the recess receives the entire pile. Socket connections are covered in the next section of the report. The following is a technology readiness form for this article. Because experimental research is limited to the tests conducted at Iowa State University (Wipf et al., 2009) and specific design specifications have not been developed, this article scores marginally above the minimum criterion for acceptance in the design specifications.

NCHRP Project 12-102 127 Figure 5.5.3.25-1 Technology Readiness Evaluation for Article 3.6.6.6

NCHRP Project 12-102 128 -10 -5 0 5 10 -400 -300 -200 -100 0 100 200 300 400 Drift [%] M om en t [ kN -m ] 5.5.3.26 Article 3.6.7.1 Socket Connections General Provisions in subsections of Article 3.6.7 are largely based on design specifications and supporting research included in report FHWA-HIF-13-037 (Marsh et al., 2013a) and also on design recommendations given by Tazarv and Saiidi (2015) and by Stephens et al., (2013). Test specimens with socket connections that were sized using current design standards and specific details included in this specification were found to emulate the seismic response of reference CIP counterparts (Motaref et al., 2011; Haraldsson et al., 2013a and 2013b; Tran et al., 2013 and 2014; Mashal et al., 2014, Mehrsoroush and Saiidi 2016; Stephens et al., 2016; Kappes et al., 2016). Figure 5.5.3.26-1 shows the measured base moment versus drift response of a column-footing assembly with wet socket connection tested at the University of Washington. A “wet socket” connection is one in which the precast column is first placed in the excavation, then fresh concrete is poured around it to create the footing. By contrast, a “dry socket” refers to a footing that is constructed with an opening, into which the precast column is subsequently placed, then grout is placed between the column and footing to form a composite assembly. Significant energy dissipation and stable hysteresis cycles were observed for drift ratios up to 6%, while only hairline cracks developed on the side of the footing. Source: Haraldsson et al., (2013) Figure 5.5.3.26-1 Drift-Versus-Column Base Moment Response for Assembly With Wet Socket Connection (Specimen SF2) Examples of implementation of socket connections in column-to-spread footing include the US 12 bridge over Interstate 5 at Grand Mound, WA. According to Belleri and Riva (2012), precast columns grouted into preformed sockets of footings have also been used extensively in Italy for commercial buildings such as warehouses and malls. The following is a technology readiness form for this article. Because socket connections have been investigated at the element and assembly levels in addition to being successfully implemented in the field, this general article scores high in technology readiness.

NCHRP Project 12-102 129 Figure 5.5.3.26-2 Technology Readiness Evaluation for Article 3.6.7.1

NCHRP Project 12-102 130 5.5.3.27 Article 3.6.7.2 Precast Concrete Column in Spread Footing or Pile Cap Socket Connection The seismic performance of socket connections for concrete column-to-spread footings have been extensively studied over the last two decades through laboratory testing and finite element modeling (Osanai et al., 1996, Motaref et al., 2011, Haraldsson et al., 2013a and 2013b, Belleri and Riva 2012, Mashal et al., 2014). Design specifications have also been developed for “wet socket” connections in which a footing is cast around a precast column (Marsh et al., 2013b). One of the key aspect to the structural performance of a socket connection, under both lateral and gravity loads, is the depth of embedment of the column into the footing. It is generally accepted, according to the related literature, that the plastic moment capacity of a column can be developed for embedment depths greater than 1.0Dc, where Dc is the diameter of the column. An experimental column- footing assembly with a shallow footing of only 0.5Dc thickness has been observed to provide a cyclic load response that was very similar to a cast-in-place assembly with the same geometry and reinforcement (Haraldsson et al., 2013a and 2013b). However, footing damage under a mechanism of combined punching shear and moment transfer was observed in that case. Because such mode of failure is not currently included in AASHTO specifications and since excessive amount of footing reinforcement is needed for a reduced lever arm, the minimum embedment depth that is proposed in the guide specification is 1.0Dc. The use of headed bars at the embedded end of the column has been found to provide a more direct load path for the transfer of moment and shear into the footing. This results in improved joint behavior as compared to the traditional bent-out reinforcement layout in CIP construction (Haraldsson et al., 2013a). Because of this, mechanical anchorages in column bars are required in the proposed guide specifications. In order to improve joint shear performance, it is also required that mechanical anchors be placed below the bottom mat of footing reinforcement for column reinforcement projected downwards. Similarly it is required that mechanical anchors be placed above the top mat of pile cap reinforcement for pile reinforcement projecting upwards. Transfer of column gravity loads in the connection is achieved through shear friction along the socket perimeter. As shown by Osani et al., (1996), for embedment depths greater than 1.5Dc transfer of vertical forces can be achieved as in CIP construction even without intentionally roughening the column’s surface. For shallower embedment depths, intentionally roughening the embedded portion of the column to minimum amplitude of 0.25 inches is proposed as a requirement to enhance shear friction. Unpublished studies suggest that less aggressive roughening techniques may also be adequate, but reliable evidence is not yet available. Tests conducted at the University of Washington (Heraldsson et al., 2013a) indicate that shrinkage of the footing concrete in a socket connection (Figure 5.5.3.27-1) results in radial compression across the interface with the column, which provides shear transfer capacity to the connection. In such case, allowing the use of a cohesion coefficient that is consistent with AASHTO LRFD Bridge Design Specifications (2014) is appropriate provided that the interface between the column and footing is clean and free from laitance. This cohesion component alone may preclude the need for additional shear friction reinforcement to support the gravity loads acting on the column as shown next:

NCHRP Project 12-102 131 Embedded ColumnCIP Footing cDc Dc Figure 5.5.3.27-1 Precast Column with CIP Footing Figure 5.5.3.27-2 Vertical Equilibrium of Embedded Portion of PC Column in CIP Footing- Cohesion Only • If the column embedment is less than 1.5Dc, then an intentionally roughened surface is required according to the proposed guide specification. For a circular column that is embedded Dc into the footing (minimum allowed), equilibrium in the vertical direction (Figure 5.5.3.26-2) gives: ܲ = ߪ గ(஽೎)మସ = ܿ(ߨܦ௖)ܦ௖ (5.5.3.27-1) where, ߪ is the axial stress in the column. For an intentionally roughened surface AASHTO LRFD Bridge Design Specification (2014) allows to use c = 0.24 ksi (less than ܭଵ ௖݂ᇱ = 0.25 ௖݂ᇱ and ܭଶ = 1.5 ݇ݏ݅) and thus Equation (5.5.3.27-1) gives: ߪ = 4ܿ = 0.96 ݇ݏ݅ (5.5.3.27-2) • If the column embedment is more than 1.5Dc, then an intentionally roughened surface is not required in the proposed specifications. Substituting the embedment depth to 1.5Dc in the free body shown Figure 5.5.3.27-2 leads to the following after establishing equilibrium in the vertical direction: ߪ = 6ܿ (5.5.3.27-3)

NCHRP Project 12-102 132 P CC T T For concrete placed against clean concrete but not intentionally roughened, AASHTO LRFD Bridge Design Specification (2014) allows to use c = 0.075 ksi (less than ܭଵ ௖݂ᇱ = 0.2 ௖݂ᇱ and ܭଶ = 0.8 ݇ݏ݅) and thus: ߪ = 0.45 ݇ݏ݅ (5.5.3.27-4) The level of axial stress due to factored vertical loads on bridge columns, would typically be less than 0.45 ksi so using the cohesion factor in AASHTO LRFD Bridge Design Specification (2014) would generally preclude the need of shear friction reinforcement in the pocket connection. In absence of cohesion, shear friction capacity through the socket of a footing can be provided by the bottom longitudinal reinforcement of the footing itself. As shown in the free-body diagram of Figure 5.5.3.27-3, equilibrium of the system under a gravity load, P, that is sufficiently large to cause yielding of the bottom steel leads to: ܲ = 2ߤ(ܥ) (5.5.3.27-5) where, C is the internal compression at footing’s interface with the column. This expression works for rectangular and circular columns. For circular columns the compression components in a given direction vary. However, the compression resultant must line up with the resisting tensile steel force from the reinforcement. This expression ensures that alignment. Figure 5.5.3.27-3 Shear Friction Capacity from Footing Bottom Reinforcement

NCHRP Project 12-102 133 But equilibrium of the footing section at the interface with the column requires that ܥ = ܣ௦ ௬݂ so the shear friction capacity provided by the footing bottom reinforcement ܣ௦ becomes: ܲ = 2ߤ(ܣ௦ ௬݂) (5.5.3.27-6) If the gravity load P is not sufficient large to yield the bottom reinforcement, then cohesion plus a smaller friction resistance associated to bending of the footing would provide the required vertical load capacity. In the tests conducted at the University of Washington, columns were subjected to axial loads as high as 3.5 times the design factored load with no sign of cracking or slip of the column relative to the footing, and there was no punching of the headed column reinforcement below the footing. (Note that the footing was elevated in the tests to permit such movement, if it occurred.) Although additional shear friction reinforcement was provided in the test footings, the stress in those bars did not exceed 5.0 ksi. Even when the column embedment was only 0.5Dc the push-through type of failure was not observed, but instead a conical punching type of failure was finally realized at a drift of nearly 9%. The following is a technology readiness form for this article. The connection has been tested extensively at the assembly level (column-footing) under both cyclic pseudo static and dynamic (shake table) loading. Design specifications for wet socket connections (footing cast around a precast column column) have already been developed and implemented for bridges in the state of Washington. Socket connections in precast footings have also been used in commercial buildings in Italy. Because of these reasons, this article scores high in technology readiness.

NCHRP Project 12-102 134 Figure 5.5.3.27-4 Technology Readiness Evaluation for Article 3.6.7.2

NCHRP Project 12-102 135 5.5.3.28 Article 3.6.7.3 Precast Concrete Column in Oversized Cast-in-Place Concrete Shaft Socket Connection This connection is an extension of the “wet socket” concept alluded in Article 3.6.7.2 but applicable to deep foundations. Provisions in this article are mostly taken from the design specifications included in report FHWA-HIF-13-037A (Marsh et al., 2013b). The latter are based on results from three large-scale column-drill-shaft assemblies tested in the laboratory (Tran et al., 2013 and 2014). The geometry of those test specimens corresponded to the minimum practical difference between diameters of the shaft and the column so as to represent the most critical condition for load transfer between the two elements. While shear friction is not required to support axial forces from the column (as this occurs through end bearing on the oversized shaft), intentionally roughening the embedded portion of the column is required in the proposed guide specifications to improve the interface transfer shear associated to the tension side of the column in bending. The required embedment of the PC column is conservatively based on a Class C splice (which no longer exists in current provisions of AASHTO LRFD) for the largest spliced bar in addition to the non- contact length between the column and shaft longitudinal reinforcement being spliced. This non-contact strut-and-tie representation assumes that anchorage forces from spliced bars occur at a 45-degree angle. The non-contact length should consider the maximum allowable horizontal deviation that could occur during placement of the column into the oversized shaft. The required embedded length has also allowance for concrete cover of both the column and shaft longitudinal reinforcement. Results from laboratory experiments by Tran et al., (2013 and 2014) indicate that if adequate confinement steel is provided around the socket (Figure 5.5.3.28-1) the plastic hinge mechanism forms in the column without incurring damage in the shaft. The proposed provisions for lateral confinement in the splice zone, generally correspond to the design recommendations given by McLean and Smith (1997): ஺ೞ೓ ௦೘ೌೣ ≥ ௞௙ೠ೗஺೗ ଶగ௙೤೟ೝ௟ೞ (5.5.3.28-1) where: Ash = area of lateral confinement steel –one leg of spiral or one leg of welded hoop (in2) smax = spacing between lateral confinement steel (in.) k = efficiency factor ful = tensile strength of column longitudinal reinforcement (ksi) Al = total area of column longitudinal reinforcement (in2) fytr = yield stress of lateral or transverse reinforcement (ksi) ls = splice length of controlling bar (in.)

NCHRP Project 12-102 136 Figure 5.5.3.28-1 Confinement Reinforcement for Column Embedded in Oversized Shaft Different values of the efficiency factor, k, are specified for the lower and upper half of the embedded portion of the splice zone. The amount of spiral in the upper half (k = 1) is twice as much as that in the lower half (k = 0.5) as damage that tends to concentrate in that region, as observed in the University of Washington experimental program. In order to provide tie reinforcement reacting against prying forces introduced by the column itself, the upper 1’-0” of the splice zone is specified to have double the amount (k = 2) as compared to the remaining portion of the upper half. At the end of the full-scale tests conducted by Murcia-Delso et al. (2016) upon removal of pieces of concrete at the top of the shaft, a cone shape surface formed between the column and the shaft steel cage at an approximate angle of 25 degrees. Since columns and shafts were 4 and 6-ft in diameter respectively, the vertical projection of this conical failure surface was about (6݂ݐ − 4݂ݐ) × tan(25଴) = 0.47 ݂ݐ. For the prototype column and shafts of the test specimens at the University of Washington (Tran et al., 2013) diameters were 6-ft and 9-ft in respectively, so the vertical projection of the expected failure cone would have been about (9݂ݐ − 6݂ݐ) × tan(25଴) = 0.70 ݂ݐ in that case. The proposed 1-ft splitting zone at the upper part of the splice is therefore judged to be conservative for practical column/shaft sizes. Experimental results from four full-scale column in oversized shafts connections conducted at the University of California San Diego (Murcia-Delso et al., 2016) led to the conclusion that the embedment length for the column reinforcement can be significantly reduced with respect to what is currently required in AASHTO Guide Specifications for LRFD Seismic Bridge Design. The minimum embedment length proposed in that study is nearly identical to Eq. (3.6.7.3.1-1) except that the Class C splice length, ls, is replaced by the tension development length ld. The resulting reduced embedment length, however, requires confinement reinforcement to meet Eq. (5.5.3.28-2): ஺ೞ೓௦೘ೌೣ ≥ ே೎೚೗ఛ೘ೌೣௗ್,೎೚೗ ଶగ௙೤೟ೝ (5.5.3.28-2) where: ௖ܰ௢௟ = number of longitudinal bars in the column ݀௕,௖௢௟ = diameter of column longitudinal bars (in.) ߬௠௔௫ = maximum bond strength of the bars, which can be taken as 2.4 ksi for ௖݂ᇱ = 5݇ݏ݅ based on bond-slip tests conducted by Murcia-Delso et al., (2013).

NCHRP Project 12-102 137 Equaling the right sides of Equations (5.5.3.28-1) and (5.5.3.28-2) and taking ߬௠௔௫ = 2.4 ݇ݏ݅, ௨݂௟ = 95 ݇ݏ݅, and setting ݈௦ = ߚ݀௕,௖௢௟, (where ߚ is a coefficient) the following relation is obtained after rearranging: ݇ = ߚ/(30ߨ) (5.5.3.28-3) This indicates that for a development length of 50 bar diameters: ߚ = ௟ೞௗ್,೎೚೗ = 1.7 × 50 Eq. (5.5.3.28-2) would produce an efficiency factor ݇ = 2.7 in Eq. (5.3.3.28-3). That means 35% more transverse reinforcement throughout the full embedment length than the amount required in the upper one foot of the connection for the proposed guide specifications. In the latter the required amount of transverse reinforcement is large in the upper one foot (where splitting cracks concentrate) and decrease away from the top. The distribution of transverse is therefore believed to be more consistent with the distribution of splitting stresses that has been observed in tests. Consider a 6ft-diameter column with 48#10 longitudinal bars (1.5% longitudinal reinforcement ratio). Assuming a development length of 50 bar diameters and ASTM A615 grade 60 reinforcement, Eq. (5.5.3.28-1) can be evaluated for k = 2 as follows: ஺ೞ೓ ௦೘ೌೣ ≥ ௞௙ೠ೗஺೗ ଶగ௙೤೟ೝ௟ೞ = ଶ×ଽହ௞௦௜×଺ଵ௜௡మ ଶగ×଺଴௞௦௜×(ଵ.଻×ହ଴×ଵ.ଶ଻௜௡) = 0.28 ௜௡మ ௜௡ , which could be satisfied with 2-#6 bundle spiral at 3.0 inches (minimum spacing). Although this is a relatively heavy confinement reinforcement, the requirement only applies to the upper 1’-0” of the embedded portion of the column. For the same conditions and assuming ௖݂ᇱ = 5݇ݏ݅, Eq. (5.5.3.28-2) on the other hand, would give: ஺ೞ೓ ௦೘ೌೣ ≥ ே೎೚೗ఛ೘ೌೣௗ್,೎೚೗ ଶగ௙೤೟ೝ = ସ଼×ଶ.ସ௞௦௜×ଵ.ଶ଻௜௡. ଶగ×଺଴௞௦௜ = 0.39 ௜௡మ ௜௡ Realizing that the required transverse reinforcement spacing in the embedded portion of the column can become unreasonable, Murcia-Delso et al., (2016) developed an expression for the minimum wall thickness of an auxiliary steel encasing ݐ௖,௠௜௡ as a function of the maximum allowable radial splitting crack width, ݑ௖௥,௠௔௫, as follows: ݐ௖,௠௜௡ = ଵఈమ௙೤,೎ ቀ ଵ ଶగ ௖ܰ௢௟߬௠௔௫݀௕,௖௢௟ − ߙଵ ஺೟ೝ ௦ ௬݂௧௥ቁ (5.5.3.28-4) where, ௬݂,௖ = nominal yield strength of steel casing, ܣݐݎ/ݏ = provided confinement reinforcement, and ߙଵ = ௨೎ೝ,೘ೌೣேೞ೓గ஽೐ೣ೟ఌ೤೟ೝ ≤ 1.0 (5.5.3.28-5) ߙଶ = ௨೎ೝ,೘ೌೣேೞ೓గ஽ೞఌ೤,೎ ≤ 1.0 (5.5.3.28-6)

NCHRP Project 12-102 138 In which ௦ܰ௛ = number of shaft longitudinal bars; ܦ௘௫௧ =diameter of shaft transverse spiral or hoops; ߝ௬௧௥ =yield strain of transverse reinforcement; ܦ௦ =diameter of steel casing; ߝ௬,௖ =yield strain of casing steel. The publication of Eq. (5.5.3.28-4) and (5.5.3.28-6) took place after the proposed guide specifications were prepared and thus they were not included in the document. When the design relies on a steel casing for confinement, allowance for corrosion should be considered and longitudinal CJP welds should be used to form the casing itself. The following is a technology readiness form for this article. Test results involve actual assemblies rather than just components. Although no record was found of implementation in the field, the connection is emulative of CIP column to oversized shaft that has commonly been constructed in WSDOT projects.

NCHRP Project 12-102 139 Figure 5.5.3.28-2 Technology Readiness Evaluation for Article 3.6.7.3

NCHRP Project 12-102 140 5.5.3.29 Article 3.6.7.4 Precast Concrete Column in Precast Pier Cap Socket Connection The seismic performance of socket connections in precast elements has been assessed through multiple monotonic cyclic load testing of scaled assemblies. In order to form the socket of column-to-pier cap connections, Mehrsoroush and Saiidi (2014) used a stay-in-place corrugated steel pipe to enhance shear friction resistance and provide joint confinement. For column-to-precast footing socket connections, on the other hand, traditional means were implemented to form the socket and this was accompanied by roughening the mating surfaces of the connecting elements (Motaref et al., 2011; Belleri and Riva, 2012; Mashal et al., 2014). As a point of clarification, a “socket” connection is one in which one member is inserted into another before grouting or placing concrete to form a composite unit. A “pocket” connection is one in which two members are placed adjacent to one another, and reinforcing steel extends from the members into a void or pocket between the two. The two members are then made composite by filling the void with grout or concrete. Experimental results by Restrepo et al., (2011) suggest that using a helical lock-seam corrugated steel pile to form a pocket in a precast pier cap provides confinement and may eliminate the need for conventional hoops in the joint. Consistent with that, it is proposed that a corrugated steel pipe always be used to form the socket in the pier cap and that the required pipe thickness should be calculated in accordance to Article 3.6.6.4. Research is needed to determine the vertical load transfer characteristics and bond behavior at the concrete-corrugated pipe interface of socket connections. Mehrsoroush and Saiidi (2014) tested column-to-pier cap assemblies in which the pier cap was provided with transverse reinforcement around the corrugated steel pipe forming the socket. Measured strains in that transverse reinforcement were as high as one third the yield strain of the material, but decreased rapidly away from the column-pier cap interface. This suggests that transverse reinforcement around the socket did have some influence on the confinement of the connection, and it is therefore conservatively specified in the proposed specifications to further maintain integrity of the joint. This provision was also recommended by Tazarav and Saiidi (2015). In the all laboratory specimens of precast elements with socket connections that were identified in the literature, the pocket depth into footing or pier cap was at least equal to the column diameter, ܦ௖. Test results suggest that such embedment depth is sufficient to develop the plastic moment capacity of the column at the interface with the connecting element. As a result, such embedment depth is established as the minimum in the proposed design guidelines, but the depth should also not be less the development length of the column reinforcement. In the commentary of the proposed specifications, end heads in the column reinforcement are strongly recommended because a more direct path to transfer loads into the connecting member and enhanced joint performance have been observed in the laboratory (Haraldsson et al., 2013a and 2013b). Other provisions regarding the detailing of the receiving precast element come from recommendations given by Tazarv and Saiidi (2015). These are based on the compilation analyzes and test results for pocket and socket connections that are available in the literature. The depth of the pier cap above the socket should be sufficiently large to avoid concrete cracking during lifting operations, and also to avoid punching shear failure due to self-weight of the pier cap and associated construction loads. The minimum top slab thickness of 0.25 times the socket length corresponds to recommendations derived from laboratory tests, although this thickness may also be controlled by handling and erection loads, in addition to the loads in the final composite condition.

NCHRP Project 12-102 141 The following is a technology readiness form for this article. Laboratory testing at the assembly level together with field implementation and developed design recommendations and provisions result in a high score for technology readiness. Figure 5.5.3.29-1 Technology Readiness Evaluation for Article 3.6.7.4

NCHRP Project 12-102 142 Annular Plate (welded) CIP FOOTING ݈௘ 8t 8t Tube Thickness (t) t CFST 5.5.3.30 Article 3.6.7.5 CFST in Cast-in-Place Concrete Footing or Pier Cap The following provisions were not included in the proposed guide specifications because AASHTO Subcommittee on Bridges and Structures had previously decided that the associated requirements and details needed further development. The connection refers to the configuration shown in Figure 5.5.3.30-1 where a steel tube is embedded into the receiving pier cap or footing and no dowel bars are involved. The detail was developed at the University of Washington following a multi-phase research project in which 19 half-scale bridge subassemblies were tested in the laboratory. In those specimens footings were detailed to resist the plastic moment capacity of the CFST. Some of the most relevant results from that research are provided by Stephens et al., (2013). Figure 5.5.3.30-1 CFST-to-CIP Footing Socket Connection In order to improve anchorage of the CFST and produce a more uniform distribution of bearing stresses, an annular steel plate (Figure 5.5.3.30-1) is welded to the embedded end of the steel tube using a complete joint penetration or fillet weld on both the inside and outside of the tube. The welds must be capable of developing the full tensile capacity of the tube, while the annular steel plate must have the same thickness and equal or greater nominal yield stress as that of the steel tube. The annular plate is specified to extend inside and outside the tube a distance of eight times the tube thickness. The embedded steel tube is to be placed above the bottom mat reinforcement of the footing or below the top mat reinforcement of the pier cap. During the experimental program, the length of embedment of the steel tube into the receiving element was found to have a significant influence on the energy dissipation capacity and mode of failure of the assembly. Figure 5.5.3.30-2 shows a comparison of the measured moment at the base of the CFST versus the measured drift ratio for two nearly identical specimens with alternative footing embedment depths. Not only were the hysteresis loops of a specimen embedded 0.6Dc narrower, indicating less energy dissipation capacity, but the mode of failure was brittle pullout of the CFST. By contrast, a specimen embedded 0.9Dc exhibited a more stable cyclic plastic hinging response and the mode of failure was tearing of the tube after several cycles of buckling and yielding.

NCHRP Project 12-102 143 Source: Stephens et al., (2013) Figure 5.5.3.30-2 Column Base Moment versus Drift Response of CFST-Footing Socket Connections with Different Embedment Depths The required embedment depth, ݈௘, of the CFST depends on whether yielding or plastic moment strength of the CFST is to be developed. For low seismic risk (such as in SDCs A and B or Seismic Zones 1 and 2), where developing the yielding moment capacity of the CFST is appropriate for design, ݈௘ is given by: ݈௘ ≥ ஽బଶ ൦ඨ1 + 16 ቀ ஽೎௧ ஽బమ ቁ ி೤ ට௙೎೑ᇲ − 1൪ (5.5.3.30-1) where: Dc = diameter of the column (in.) t = thickness of annular ring (in.) D0 = Dc +16t = outside diameter annular ring (in.) Fy = nominal yield stress of the tube (ksi) ௖݂௙ᇱ = compressive strength of footing or pier cap concrete (ksi) For high seismic risk (such as in SDCs C and D or Seismic Zones 3 and 4) where developing the plastic moment capacity is appropriate for design, ݈௘ is given by: ݈௘ ≥ ஽బଶ ൦ඨ1 + 21 ቀ ஽೎௧ ஽బమ ቁ ிೠ ට௙೎೑ᇲ − 1൪ (5.5.3.30-2) where: Fu = expected tensile strength of the tube (ksi) Equation (5.5.3.30-2) was derived by Stephens et al., (2013) assuming that the tensile strength of the tube is developed simultaneously to concrete pullout at a shear stress of

NCHRP Project 12-102 144 Plastic Neutral Axis Fy Fy 0.95f’c Concrete Stress Steel Stress Equilibrium Acc Asc Ast Cc =0.95f’cAcc Cs =FyAsc Ts =FyAst Py ݒ௡ = 0.19ට ௖݂௙ᇱ (in ksi units) as illustrated in Figure 5.5.3.30-3. This limiting unit strength of concrete was obtained from calibrating a simplified pullout model with the test results. Figure 5.5.3.30-3 Determination of Flexural Capacity of CFST Using the Plastic Stress Distribution Method From the research conducted at the University of Washington it was also specified that the concrete depth beyond the tube must be at least 12 times the steel tube thickness or 6 inches and satisfy Equation (5.5.3.30-3) to prevent the occurrence of punching shear failure: ݐ௙ ≥ ඨ ஽೎మ ସ + ହ.ଷ(஼೎ା஼ೞ) ට௙೎೑ᇲ − ஽೎ଶ − ݈௘ (5.5.3.30-3) where: Dc = diameter of the column (in.) ௖݂௙ᇱ = compressive strength of footing or pier cap concrete (ksi) Cc = compressive force in the concrete due to combined axial load and moment as calculated from using the plastic stress distribution method (kip) Cs = compressive force in the steel due to combined axial load and moment as calculated from using the plastic stress distribution method (kip) The plastic stress distribution method, illustrated in Figure 5.5.3.30-4, was found to provide a reasonable estimate of the plastic moment capacity of the CFST when the steel tube is properly anchored into the receiving member.

NCHRP Project 12-102 145 45o Figure 5.5.3.30-4 Simplified Pullout Model for CSFT Figure 5.5.3.30-5 Reinforcement Detail at CFST-to-Footing Connection Detailing requirements for load transfer across the joint of the CFST socket connection (Figure 5.5.3.30-5) are not explicitly provided by Stephens et al. (2013). WSDOT Bridge Design Manual (2016), however, presents this configuration as a preferred “fully restrained” connection for CFST columns and provides the following specific detailing requirements: • The required edge distance, ݀௘, from the center of the CFST to the edge of the footing or pier cap must satisfy: ݀௘ ≥ ܦ௖ (5.5.3.30-4) This provision is based on recommendations by Stephens et al., (2016) following additional analytical and experimental research of CFST socket connections in precast pier caps. From that

NCHRP Project 12-102 146 study it was concluded that a properly detailed pier cap with a width as narrow as 2.0 times the tube diameter is sufficient to develop the plastic moment capacity of the CFST. • Vertical ties must be placed in the region of the connection within a distance 1.5݈௘ from the outside of the CFST at spacing ݏ satisfying: ݏ ≤ ௟೐ଶ.ହ (5.5.3.30-5) This provision is intended to ensure that the assumed pullout failure cone is crossed by multiple vertical ties. • It is assumed that concrete provides one third of the unit shear strength ݒ௡ (= 0.19ට ௖݂௙ᇱ ), across the pullout plane. Although not explicitly stated in the WSDOT Bridge Design Manual (2016), this implies that vertical ties could be specified to satisfy Equation (5.5.3.30-6) within a horizontal distance ݈௘ so as to provide the remaining part of the unit shear capacity (i.e., 2/3 × ݒ௡ = 0.13ට ௖݂௙ᇱ ). ߩ௩ ≥ ଴.ଵଷට௙೎೑ᇲ ௙೤೟ (5.5.3.30-6) where: ௬݂௧ = nominal yield stress of vertical tie reinforcement (ksi). ߩ௩ = ஺ೡ,೟೔೐௦௕ = vertical tie reinforcement ratio ܣ௩,௧௜௘ = tie cross-sectional area (in2) ܾ = width tributary to a tie (in.) Following this approach the total amount of vertical ties that is required around the CFST is linearly proportional to the embedment length. This means that stronger and thicker steel tubes would not only need more embedment depth but also more vertical ties, as may be expected. It is clear, however, that detailing requirements for design have not been fully developed for the CSFT socket connection. One consideration is tying the amount of joint shear reinforcement to the amount of steel developed in the plastic mechanism but increased to account for an overstrength factor (similar to the methods outlined in AASHTO Guide Specifications for LRFD Seismic Bridge Design). This can be evaluated using fully developed free-body diagrams to ensure the forces are transferred correctly through the connection.

NCHRP Project 12-102 147 5.5.3.31 Article 3.6.8 Full-Depth Precast Concrete Deck Panel Connections Section 4.2.1 of this report describes in detail the various research projects that have been undertaken regarding full-depth precast concrete deck panels. The research to date on precast concrete deck panels is more significant than any other type of prefabricated element. The basis of the provisions for the Guide Specification is the PCI State of the Art Report on Full-Depth Precast Concrete Bridge Deck Panels (2011). This report includes design provisions and sample calculations. In general, the following approach is used for the design of full-depth precast concrete deck panel connections: • The design of the reinforcing within the panel connections can be based on standard cast-in-place reinforced concrete deck design in the AASHTO LRFD Bridge Design Specifications. The “strip method” of deck design in applicable to full-depth precast panels. • Secondary (distribution) connections can be designed according to the provisions in the AASHTO LRFD Bridge Design Specifications for a cast-in-place reinforced concrete deck. • Post-tensioning may be used for joints in the secondary (distribution) direction. The design should follow the provisions for these joints in the AASHTO LRFD Bridge Design Specifications. • Composite action connections can be achieved through the use of reinforced dowel pockets. The design of the dowels or shear connectors can be based on the provisions for cast-in-place concrete in the AASHTO LRFD Bridge Design Specifications. The following is a technology readiness form for this article. The vast body of existing research and common use of these elements result in a high score for technology readiness.

NCHRP Project 12-102 148 Figure 5.5.3.31-1 Technology Readiness Evaluation for Article 3.6.8

NCHRP Project 12-102 149 5.5.3.32 Article 3.6.9 Link Slabs Section 4.2.1 of this report describes in detail the various research projects that have been undertaken regarding link slabs. The basis of the provisions for the Guide Specification is the PCI Journal article entitled “Behavior and Design of Link Slab for Jointless Bridge Decks (Caner, et al., 1998)”. This article includes design provisions and the theory of link slabs. The general approach for link slab design is to accommodate girder end rotation within a debonded portion of the deck end. The recommended portion of debonded deck is equal to five percent of each span. Once rotations are calculated, a design moment is calculated based on a simple free-body diagram of the slab. The provisions in the guide specifications include the assumed free-body diagram and a detail showing the debond zone. The following is a technology readiness form for this article. The body of existing research and common use of these elements result in a high score for technology readiness.

NCHRP Project 12-102 150 Figure 5.5.3.32-1 Technology Readiness Evaluation for Article 3.6.9

NCHRP Project 12-102 151 5.5.3.33 Article 3.6.10 Steel Connections At this time, there are no special methods for connecting steel elements. Connections are made with either bolting or welding. Bolted connections can be time consuming, therefore this provision includes recommendations on how to eliminate bolted field splices in girders. The use of span-by-span construction is recommended. Joints should be avoided in bridge decks in order to improve durability, therefore the use of link slabs are recommended for this scenario. The concept of designing span-by-span construction using the “simple san for dead load, continuous for live load” is included in this section. This approach relies on a moment connection at the interior piers after the casting of the deck. Research that has been completed on this connection has been referenced for use by designers that desire to employ this method. A technology readiness evaluation was not completed for this article since the technologies listed are covered by the AASHTO LRFD Bridge Design Specifications. 5.5.3.34 Article 3.6.11 Integral Substructure Connections No specific provisions are provided for this article, since general engineering principals are used to make these connections. The designer should calculate the forces acting on the connection by applying the appropriate loads to the structural system.

NCHRP Project 12-102 152 5.5.3.35 Article 3.6.12 Two-Stage Integral Pier Cap Figure 5.5.3.35-1 illustrates the two-stage integral pier cap system. This article refers to the design of a force transfer mechanism between the dropped cap beam and the upper superstructure diaphragm. Such configurations are often used to simplify construction by providing a lower-stage cap on which girders may be supported. Later, the upper stage of the cap is completed by placing concrete above the lower stage. See Figure 5.5.3.35-1 for the construction sequence. Adequate detailing needs to be provided through the interface between the two components to ensure composite action. Because the joint shear force transfer between the vertical seismic system and superstructure takes place in the upper pier cap, the forces from the column longitudinal reinforcement may be transferred into the upper cap by extending the column bars, by using the vertical legs of tie reinforcement crossing the interface, or by other means. The integral dropped cap has the effect of enlarging the effective joint shear region by virtue of increasing the torsional stiffness of the joint itself. For lateral loads in the longitudinal direction of the bridge, joint shear force transfer occurs in the upper cap only. For lateral loads in the transverse direction of the bridge, on the other hand, shear force transfer occurs in the full two-stage integral pier cap. Because of the two-stage configuration, failure modes that are not necessarily present with a flush- soffit (single stage) cap must be considered. For example, stirrups in the two-stage cap may be required to handle not only shear, but also participate in the transfer of moment from the column through the lower- stage cap to the upper stage. Torsional force transfer along the composite cap will also be different in the two-stage cap configuration. Rational strut-and-tie models can be helpful in ensuring that all possible failure modes are considered. The following is a technology readiness form for this article. The two-stage integral cap system is widely used and provisions to ensure composite behavior between components of an element are well developed and known. As a result, this item scores high for technology readiness. Figure 5.5.3.35-1 Construction Sequence of Two-Stage Integral Pier Cap System

NCHRP Project 12-102 153 Figure 5.5.3.35-2 Technology Readiness Evaluation for Article 3.6.12

NCHRP Project 12-102 154 5.5.3.36 Article 3.6.12.1 Joint Proportioning for Two-Stage Integral Pier Cap for SCDs C and D or Seismic Zones 3 and 4 The requirements in this article are nearly identical to the “Joint Design for SDCS C and D” provisions outlined in AASHTO Guide Specifications for LRFD Seismic Bridge Design (1) with the exception that the provisions are extended to force-based design in Seismic Zones 3 and 4. In addition, conditions and substitute equations are provided to define the joint geometry in a manner that is consistent with the load path in each direction of analysis as summarized next: • Longitudinal Direction: For the calculation of principal stresses in the joint region under plastic hinging moment applied perpendicular to the cap (applied in the bridge longitudinal direction with no skew), the depth of the joint is reduced to the depth of the superstructure, Ds2, while the effective width of the joint, Beff, increases based on an assumed 45-degree angle surface that starts at the column face and extends to the mid-depth of the superstructure. The same effective width is used for the calculation of the average vertical stress, ௩݂. The wider width for stress calculations is justified by the ability of the dropped beam to distribute forces along its length prior to reaching the longitudinal joint region. In no case Beff shall be taken more than the width that is geometrically tributary to the column. Similarly, for the calculation of the average joint vertical shear stress, ݒ௝௩, only the length of the reinforcement, ݁௨௖, extending from the transverse cap and column into the superstructure is considered. Modified joint average stress equations are proposed in the specifications to account for these dimensions and the corresponding effective stress planes. Figures 5.5.3.36-1 through 5.5.3.36- 3 show the proposed critical areas for the calculation of joint stresses. • Transverse Direction: For the calculation of principal stresses in the joint region under plastic moment applied parallel to the cap (applied in the transverse direction with no skew), the depth of the joint is equal to the combined depth of the two-stage integral pier cap, ܦ௦ଵ + ܦ௦ଶ. Critical areas for the calculation of ௛݂, ௩݂, and ݒ௝௩ are shown in Figures 5.5.3.36-4 through 5.5.3.36-6 respectively. Figure 5.5.3.36-1 Critical Area for the Calculation of fh for Seismic Excitation in the Longitudinal Direction of the Bridge

NCHRP Project 12-102 155 Figure 5.5.3.36-2 Critical Area for the Calculation of fv for Seismic Excitation in the Longitudinal Direction of the Bridge Figure 5.5.3.36-3 Critical Area for the Calculation of vjv for Seismic Excitation in the Longitudinal Direction of the Bridge

NCHRP Project 12-102 156 Figure 5.5.3.36-4 Critical Area for the Calculation of fh for Seismic Excitation in the Transverse Direction of the Bridge Figure 5.5.3.36-5 Critical Area for the Calculation of fv for Seismic Excitation in the Transverse Direction of the Bridge Ds2 Ds1 Dc Bcap

NCHRP Project 12-102 157 Figure 5.5.3.36-6 Critical Area for the Calculation of vjv for Seismic Excitation in the Transverse Direction of the Bridge The following is a technology readiness form for this article. A high score is assigned because the system is widely used and because only modifications to existing AASHTO joint average stress equations are being proposed for consistency with the load paths in each direction of analysis. A reduction in the resulting scoring index manifests the insufficient research and laboratory testing on the seismic behavior of dropped cap beam system with integral connection. These provisions had already been proposed in FHWA Report No. HIF-13037-A (2). euc Ds1 Dc Beff

NCHRP Project 12-102 158 Figure 5.5.3.36-7 Technology Readiness Evaluation for Article 3.6.12.1

NCHRP Project 12-102 159 5.5.3.37 Article 3.6.12.2 Minimum Joint Shear Reinforcing for SDCs C and D or Seismic Zones 3 and 4 The requirements in this article are nearly identical to the “Minimum Joint Shear Reinforcing” provisions outlined in AASHTO Guide Specifications for LRFD Seismic Bridge Design (1) with the exception that for joints in which the calculated principal tension exceeds the cracking limit of 0.11ඥ ௖݂ᇱ the designer should select the maximum transverse reinforcement ratio (in the horizontal plane) between the one required for an uncracked joint and that required for a cracked joint. Restrepo et al. (2011) have pointed out that there is no guarantee that either reinforcement ratio is larger than the other. The following is a technology readiness form for this article. A high score is assigned because the provisions already exist and only a minor and conservative modification has been introduced to provide joint shear reinforcement.

NCHRP Project 12-102 160 Figure 5.5.3.37-1 Technology Readiness Evaluation for Article 3.6.12.2

NCHRP Project 12-102 161 5.5.3.38 Article 3.6.12.3 Superstructure Capacity Design for Two-Stage Integral Pier Caps for Longitudinal Direction for SDCs C and D The requirements in this article are nearly identical to the “Superstructure Capacity Design for Integral Bent Caps for Longitudinal Direction for SDCs C and D” provisions outlined in AASHTO Guide Specifications for LRFD Seismic Bridge Design (1) with the exception that the effective width of the superstructure resisting longitudinal seismic moments is revised, as shown in Figure 5.5.3.38-1, for open soffit girder-deck superstructures with integral dropped cap beam. The proposed effective width accounts for the additional lateral distribution of longitudinal moments by virtue of the additional cap beam depth and associated increased torsional capacity of the composite element. The two stages of cap construction must be integral over the full width of the cap, and closed torsional stirrups and longitudinal steel must be present in the cap beam to distribute the induced torsional forces. Figure 5.5.3.38-1 Effective Superstructure Width in the Longitudinal Direction for Dropped Cap System These provisions had already been proposed in FHWA Report No. HIF-13037-A (2). The following is a technology readiness form for this article. A high score is assigned because the system is widely used and because only modifications to existing AASHTO substructure capacity design provisions are being proposed for consistency with the load paths in the longitudinal direction of analysis. A reduction in the resulting scoring index manifests the insufficient research and laboratory testing on the seismic behavior of dropped cap beam system with integral connection and its implication in superstructure capacity design.

NCHRP Project 12-102 162 Figure 5.5.3.38-2 Technology Readiness Evaluation for Article 3.6.12.3

NCHRP Project 12-102 163 5.6 Section 4 – Detailing Requirements This section includes recommendations for general detailing requirements for ABC designs using prefabricated elements. There are several sources for the provisions in this section: 1. The results of the questionnaire survey: a. Many agencies submitted plans from ABC projects b. Several agencies have bridge manual provisions and/or standard details developed 2. NCHRP Project 12-98 includes the development of provisions for tolerance management. Fortunately, Michael Culmo, the PI for that project, is the same as for this project. 3. The project team has significant experience with the development of actual plans for ABC projects. 4. The detailing requirements for GRS/IBS were taken from the FHWA Report entitled Geosynthetic Reinforced Soil Integrated Bridge System Interim Implementation Guide (2012).

NCHRP Project 12-102 164 5.7 Section 5 – Durability of ABC Technologies This section includes recommendations for detailing and design for durability. There are several sources for the provisions in this section: 1. The results of the questionnaire survey: a. Many agencies submitted plans from ABC projects b. Several agencies have bridge manual provisions and/or standard details developed 2. The Utah DOT has written several reports on the durability of ABC projects. Michael Culmo, the PI for this project was the author of those reports. The reports identify problems with certain details and features, and successes of other details. The results of the Utah DOT reports are accounted for in the provisions contained in this section. 3. The project team has significant experience with the development of actual plans for ABC projects. 4. The PCI Northeast Bridge Technical Committee has been studying ABC detailing for many years. Michael Culmo, the PI for this project has been a member of this committee for approximately 25 years. This committee, which is made up primarily of DOT engineers has always had a focus on durability in design. Items such as providing adequate cover over mechanical connectors and headed bars are taken from work of this committee. 5. Requirements of corrosion protection of post-tensioning systems was taken from FHWA and industry publications.