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28 Chapter 4 RESEARCH BASIS SUPPORTING THE PROPOSED REVISIONS TO INCORPORATE SV INTO DESIGN PROCEDURES The outcomes from NCHRP Project 24-1 research study that provide justification for the recommended revisions to the AASHTO Design Method presented in Chapter 3 are provided herein. Consistent with Chapter 3, the research basis supporting the proposed revisions into current AASHTO design procedures are also grouped into five design aspects. In addition, an initial discussion is provided on the identification of the magnitude of reinforcement vertical spacing that defines the âboundaryâ of composite behavior. 4.1 BOUNDARY FOR COMPOSITE BEHAVIOR OF GEOSYNTHETIC-REINFORCED SOIL STRUCTURES Before discussing the research basis for recommended revisions on each design aspect presented in Chapter 3, an overarching important question is âwhat constitutes closely-spaced reinforcement?â, i.e., what is the vertical reinforcement spacing, Sv, below which the influence among adjacent reinforcement layers (i.e. composite behavior) should be considered. As discussed in NCHRP Project 24-41, Final Report, Section 5, Figure 4.1 shows a conceptual relationship for the interaction between neighboring reinforcement layers (i.e., interlayer interaction) and the reinforcement spacing. In this figure, a vertical spacing Sv,c (i.e. composite behavior) is indicated below which full interlayer interaction (or composite behavior) occurs. The figure also identifies a vertical spacing Sv,nc (i.e. non-composite behavior) beyond which no interlayer interaction occurs. Varying degrees of interlayer interaction develop for vertical spacing values ranging from Sv,c to Sv,nc . Figure 4.2 shows experimental data used to quantify the interaction between neighboring reinforcement layers at various reinforcement spacings. The indicator for this interaction adopted in this evaluation is the ratio between the reinforcement displacements measured in a neighbor reinforcement, v, to that measured in the active reinforcement, u. Specifically, Figures 8.4.2 a, b, and c show the relationship for active reinforcement displacements of 2, 5, and 10 mm (0.1, 0.2 and 0.4 in), respectively (i.e. three different loading tensile levels). The results in the figure indicate that full interlayer interaction occurs for vertical spacing values below approximately 0.1 m (4 in.) ((i.e. Sv,c = 0.1 m (4 in.)). On the other hand no composite behavior is considered for vertical spacing values above approximately 0.1 m (4 in.) ((i.e. Sv,nc = 0.2 m (8 in.)), as beyond this distance the interaction has been significantly reduced. Consequently, according to these experimental results, the composite behavior can be observed to an average of 0.15 m (6 in.) from the soilâgeosynthetic interface.
29 Figure 4.1. Conceptual relationship for the interaction between neighboring reinforcement layers (i.e., interlayer interaction) and the reinforcement spacing.
30 Figure 4.2. Displacement ratio versus reinforcement spacing at various active reinforcement displacements: (a) 2 mm (0.08 in); (b) 5 mm (0.2 in); and (c) 10 mm (0.4 in). (a) (b) (c) 0.00 0.05 0.10 0.15 0.20 0.25 0.30 0.35 0.00 0.05 0.10 0.15 0.20 0.25 0.30 0.35 0.40 0.45 Di sp lac em en t R ati o, v/ u Reinforcement Vertical Spacing, Sv (m) 50 kPa 35 kPa 21 kPa 15 kPa uav = 2 mm 0.00 0.05 0.10 0.15 0.20 0.25 0.30 0.35 0.00 0.10 0.20 0.30 0.40 0.50 0.60 Di sp lac em en t R ati o, v/ u Reinforcement Vertical Spacing, Sv (m) 50 kPa 35 kPa 21 kPa 15 kPa uav = 5 mm 0.00 0.05 0.10 0.15 0.20 0.25 0.30 0.35 0.00 0.10 0.20 0.30 0.40 0.50 0.60 Di sp lac em en t R ati o, v/ u Reinforcement Vertical Spacing, Sv (m) 50 kPa 35 kPa 21 kPa 15 kPa uav = 10 mm
31 The reinforcement spacing in these figures at which a significant change in the interaction between contiguous reinforcements occur corresponds to the experimentally-defined boundary for composite behavior of a geosynthetic-reinforced soil mass, Sv,c. However, this value represents a lower bound, as load in the experimental testing setup was mobilized in only one of the reinforcements. In case of multiple loaded reinforcements, the soil between two contiguous reinforcements would be mobilized in shear by the two reinforcement layers. Consequently, according to these experimental results, a composite behavior would be observed for vertical spacing values having an average of 0.30 m (12 in.) for select backfill. This value is well in agreement with current limits for reinforcement spacing established by the FHWA GRS-IBS design approach, as indicated in Table 2.1. In order to avoid having a sharp change of response at a given vertical spacing, and consistent with the experimental data shown in Figure 3.2, which indicates a transition zone between the responses at widely- and closely-spaced reinforcements, a transition is proposed for the determination of Tmax as a function of vertical spacing. Accordingly, a magnitude Sv,c of 0.2 m (8 in.). was identified as the spacing below which a composite behavior is considered, while a magnitude Sv,nc of 0.4 m (16 in.) was identified as the spacing above which a non-composite behavior shall be considered. A transition zone was proposed for reinforcement vertical spacing values ranging from Sv,c and Sv,nc (i.e. between 0.2 and 0.4 m (8 and 16 in.)). As previously mentioned, a combination of experimental results and numerical parametric evaluations showed little to no effect of the backfill strength properties on the extent of the soilâ geosynthetic interaction influence zone. Consequently, the boundary of the composite behavior of geosynthetic-reinforced soil structures is recommended for select backfill where the soilâ geosynthetic coefficient of interaction exceeds 0.8 (i.e., F* > 0.8 tan Ï). 4.2 EFFECT OF SV ON TMAX MAGNITUDE AND DISTRIBUTION In order to provide the basis for the recommended effect of Sv on Tmax magnitude and distribution, as presented in Chapter 3.2, the different research outcomes, including those from experimental, field and numerical components were holistically assessed. They are summarized next. 4.2.1 EXPERIMENTAL RESEARCH OUTCOMES THAT SUPPORT THE PROPOSED RECOMMENDATIONS The results from the experimental testing program documented in Chapter 5 of this report provide evidence on the existence of a relationship between the maximum reinforcement unit tension and the reinforcement vertical spacing. The impact of vertical spacing could be experimentally identified for vertical spacing values ranging from 0.10 m to 0.20 m (4 to 8 in) and being somewhat independent of the normal stress for ranges that are typical for working stress conditions (15 to 50 kPa (2.2 to 7.3 psi)). While load in the experimental testing program was mobilized in only one of the neighboring reinforcements, the thickness of the soil layer affected by the shear transfer would double for the case of load mobilized in multiple reinforcements. As discussed in Section 8.4.1, the center of the transition zone is 0.15 m (6 in) from each soilâ
32 geosynthetic interface (or 0.30 m (1 ft) for a soil layer bound by two geosynthetics). The design recommendations proposed in Chapter 3.2 consider full interaction (i.e. composite behavior) for the reinforcement vertical spacings below 0.40 m (8 in.). Also consistent with the experimental and numerical results, a transition between composite and non-composite behaviors was recommended in the determination of Tmax for vertical spacing values ranging from 0.40 m (8 in.) and 0.80 m (16 in.). As discussed in Section 5 of the NCHRP Project 24-41 Final Report, load shedding was observed in the experimental results for tests where the the reinforcement vertical spacing Sv was within the values recommended for implementation in AASHTO. That is, load was observed to transfer through shear bands from highly stressed geosynthetic layers to the adjacent layers that were comparatively less stressed. Ultimately, geosynthetic layers are expected to be tensioned uniformly in a reinforced soil mass, provided that the reinforcement vertical spacing is small enough to facilitate load transfer through the soil by shear. This observation implies that for conditions involving reasonably highly mobilized loads and closely-spaced reinforcements, each layer would be expected to carry a similar Tmax (assuming that all layers are long enough). 4.2.2 FIELD RESEARCH OUTCOMES THAT SUPPORT THE PROPOSED RECOMMENDATIONS Data compiled for Tmax for the GRS-IBS bridge instrumented abutment presented in Section 6 of the NCHRP Project 24-41 Final Report was plotted along with additional data compiled for other GMSE structures of different vertical reinforcement spacings presented in Section 4.6 of the NCHRP Project 24-41 Final Report. Figures 4.3a through 4.3d present measured Tmax profiles from self-weight loading condition together with the design load distribution proposed in the NCHRP Project 24-41study for four cases: Virginia Bridge, Louisiana Bridge, Delaware Bridge, and Founders Meadows Bridge. It can be observed that the distribution of Tmax with depth is reasonably uniform and does not tend to either increase with depth. In addition, Figures 4.4a through 4.4d present measured Tmax profiles from self-weight and surcharge loading condition for Virginia Bridge, Louisiana Bridge, Delaware Bridge, and Havana Yard Abutment (Wu et al. 2001 - FHWA-RD-00-038 Report). Also in this case, the distribution of Tmax with depth is reasonably uniform and does not tend to either increase with depth.
33 (a) (b) (c) (d) Figure 4.3. Measured Tmax profiles from self-weight loading condition and proposed design envelopes: (a) Virginia Bridge; (b) Louisiana Bridge; (c) Delaware Bridge â East Abutment; and (d) Delaware Bridge â West Abutment.
34 (a) (b) (c) (d) Figure 4.4. Measured Tmax profiles from self-weight and surcharge loading condition and proposed design envelopes: (a) Virginia Bridge; (b) Louisiana Bridge; (c) Delaware Bridge â East Abutment; and (d) Delaware Bridge â West Abutment.
35 4.2.3 NUMERICAL SIMULATION OUTCOMES THAT SUPPORT THE PROPOSED RECOMMENDATIONS In order to complement observations from field monitoring data, which may show significant scatter, numerical simulations were used to extend the evaluation of the effect of Sv on the magnitude and distribution of Tmax. As discussed in detail in Chapter 7 of this report, Figure 4.5 shows the effect of the reinforcement spacing on Tmax with depth, under both self-weight and traffic load conditions. It should be noted that these results represent working stress conditions as the numerical model was calibrated for structures under these conditions. The prediction of Ïv under different loading conditions was evaluated using the overburden stress and the AASHTO 2 to 1 truncation method. In Figure 4.5, z is the depth measured from top of the wall, normalized against the wall height H . The numerical predictions in Figure 4.5 shows that for self-standing walls, the numerical predictions for Tmax could not reproduce the uniform distribution observed in field structures. However, both under traffic load and under bridge load surcharge, the predicted Tmax was found to be reasonably uniform. It should also be noted that the predicted max(Tmax) is smaller than the value that would have been used for the case of structures with widely-spaced reinforcement. Figure 4.5. Effect of Sv on Tmax vs depth for: (a) Self-weight; (b) Traffic load with consideration of surcharge q; (c) Traffic load without consideration of surcharge q. 4.2.4 LIMIT EQUILIBRIUM RESEARCH OUTCOMES THAT SUPPORT THE PROPOSED RECOMMENDATIONS The LE-based methodology presented in Leshchinsky et al. (2016) Report No. FHWA-HIF-17-004, utilizing program ReSSA+, shows that when the reinforcement is long enough, Tmax among reinforcement layers is nearly uniformly mobilized. This finding results from examination of
36 numerous slip surfaces, several of which yield the same Tmax although the surfaces are different. For example, some surfaces reflect a compound mode of failure while others are strictly internal but emerging either at or above the toe of the wall. The nearly uniform mobilization of reinforcement was demonstrated analyzing a wall with a height of 6 m (20 ft), reinforcement length of 4.2 m (14 ft), and reinforcement spacing of either 0.4 m or 0.2 m (16 in. or 8 in.). Figures 4.6a and 4.6b show Tmax profile obtained from the LE analysis and the current AASHTO design envelope for zero surcharge and for 200-kPa (4000-psf) surcharge, respectively. (a) (b) Figure 4.6. Comparison between the limit equilibrium computed Tmax envelope and current AASHTO design envelope: (a) surcharge, q = 0; (b) surcharge, q = 20 kPa (430 psf). 4.3 EFFECT OF SV ON T0 MAGNITUDE AND DISTRIBUTION In order to provide the basis for the recommended effect of Sv on the magnitude and distribution of the connection load, T0, as presented in Chapter 3.3, the different research outcomes were holistically assessed. They are presented next.
37 4.3.1 FIELD RESEARCH OUTCOMES THAT SUPPORT THE PROPOSED RECOMMENDATIONS The current design models for evaluating T0 are significantly different As previously indicated in Chapter 2.2 and Table 2.1, current design models differ significantly on their evaluation of T0. Specifically, AASHTO requires T0 = Tmax, while the FHWA GRS-IBS design approach assumes that the lateral pressure on the facing is comparatively small and does not require a design. That is, friction connection is deemed to be sufficient to take on the small tensile load that the FHWA GRS-IBS approach assumes at the connection. Field data on connection loads from three different studies, as compiled by Wu and Ooi (2015) Report No. FHWA-HRT-14-094 were reassessed as part of the study in order to validate the hypothesis that the magnitude of connection loads is comparatively minor. The field data involved four piers (Mitchell 2002), the Founders Meadows load-carrying GMSE bridge abutment and two GMSE structures (Yogarajah and Andrawes 1994). Figure 4.7 compares the measured magnitude of the connection loads against predicted values using the approaches proposed by Wu (2001) and Soong and Koerner (1997), both of which have been reported to predict connection load values that are below those adopted by AASHTO. The reassessment of field data showed, however, that the measured connection loads were comparatively high, at least in relation to those predicted by the methodologies proposed by Wu (2001) and Soong and Koerner (1997). In addition, Figure 4.8 shows an increase in bias with increasing lateral earth pressure, resulting in differences of up to 15 times. This lateral pressure is expected to be larger in load-carrying GMSE bridge abutments, where a footing or bearing seat is subjected to high loads (e.g., 200 kPa (4,200 psf)), which are often expected to be near the wall facing. Figure 4.7. Predicted versus measured lateral stress against facing blocks.
38 Figure 4.8. Bias in the lateral stress prediction models. The reason for the scatter shown in Figures 4.7 and 4.8 may be the product of a number of complicating factors in the monitored structures, including lateral movement and tilting of the facing blocks as well as parasitic loads that are often induced at the facing connection due to reinforcement downdrag. Even some mini-pier tests indicated downdrag stresses on the connection, to the point that failure of the reinforcement at the connection occurred in some tests or at least very high deformation observed at the connection during dismantling (Iwamoto, 2014). In summary, a number of mechanisms may affect the magnitude of the measured connection loads, including the development of relative movements between the facing and the reinforced soil mass, the method of instrumentation (e.g., earth pressure cells, strain gages, extensometers mounted on the reinforcements), and the location of the monitoring instruments. Consequently, and in spite of reasonable scatter in the field data, the available field measurements have indicated that the magnitude of the connection loads is not necessarily negligible, which is consistent with the recommended revisions presented in Chapter 3.3. The distribution of connection loads with depth was also evaluated as part of the NCHRP Project 24-41 study. This involved reassessment of data collected as part of the field research component documented in Chapter 6 and field data collected by others as reported in Section 4.6 of the NCHRP Project 24-41 Final Report. Even though no changes are recommended in the NCHRP Project 24-41 study regarding the magnitude of T0 in relation to Tmax, changes are proposed to the AASHTO design approach regarding the distribution of Tmax with depth. These changes effectively result in a change on the distribution of T0 with depth for the case of GMSE walls with 0 5 10 15 20 25 30 0 5 10 15 20 25 30 Bi as (M ea su re d/ Pr ed ic te d) Lateral Stress (kPa) Wu (2001) Method Soong and Koerner (1997) Method Rankine's Theory Values predicted using Rankine's Theory did not account for surcharge loading Bias = 1
39 closely-spaced reinforcement. Specifically, from 3.3 and 3.5, the connection load T0 of a given reinforcement layer i can be expressed as follows: , = â â â + â â for â¤ , = 8" [Equation 4-1] Also, for reinforcement vertical spacings ranging from Sv,c and Sv,nc (i.e., from 0.2 to 0.4 m (8 to 16 in.)), the magnitude of T0,i would be defined from Equations 4.4 and 4.5, as follows: , = â â â + , , â + âÏ â S for 8" = , â¤ â¤ , = 16â [Equation 4.2] Figure 4.9 shows the profile of normalized T0 versus depth, defined using field data corresponding to strain gages (SG) and rectangular pressure cells installed in the Virginia (VA) GRS-IBS structure, SG installed in the Delaware (DE) GMSE structure, earth pressure cells (EPC) installed in the Colorado (CO) GRS-IBS structure and SG installed in the Founders Meadows (FM) load-carrying GMSE abutment. The vertical reinforcement spacing (Sv) in the VA and DE structures was 0.20 m (8 in.) in the primary reinforcement zone and 0.10 m (4 in.) in the bearing seat zone. The vertical reinforcement spacing (Sv) in the Colorado GRS-IBS and FM GMSE abutments were 0.10 and 0.40 m (4 and 16 in.), respectively. As shown in the results presented in the figure, the profile of connection loads (or its normalized value) can be reasonably described as having a uniform distribution with depth.
40 (a) (b) (c) (d) Figure 4.9. Profiles of connection load with depth for (a) the Virginia GRS-IBS structure, (b) the Louisiana GRS-IBS, (c) the Delaware GRS-IBS structure (East Abutment), and (d) the Delaware GRS-IBS structure (East Abutment).
41 Additional evidence of the appropriateness of adopting a uniform distribution with depth for the connection load, T0, is provided by the connection loads predicted at the FM load-carrying GMSE bridge abutment. As discussed in Section 4.5 of the NCHPR Project 24-41 Final Report, the reinforcement vertical spacing for this structure was 0.40 m (16 in.), which is in the upper limit of the range recommended in Chapter 3 for âclosely-spacedâ reinforcement. Figure 4.10 shows the connection loads at geogrid layers 6 and 10 placed 3.05 and 1.43 m (10 and 4.7 ft) below the bridge footing, respectively. The change in connection load is provided as a function of the vertical pressure predicted along the different stages of construction of the bridge structure, including field placement, placement of the footing and placement of the bridge girders. The connection loads were obtained using measurements from SG placed on the geogrids in close proximity to the facing blocks. As shown in the figures, the connection loads at the two different elevations remained approximately uniform, at T0 = 6 kN/m (Â± 1 kN/m) (34.2 lb/in Â± 5.7 lb/in), throughout the different stages of construction. The linear prediction that would be obtained using current AASHTO design approach would deviate significantly from the actual connection loads. Additional evidence on the proposed distribution of reinforcement connection loads with depth is provided by Leshchinsky et al. (2010) who reported observations from an exhumed geogrid-reinforced retaining wall with a vertical reinforcement spacing of 0.40 m (16 in.) constructed on highly compressible foundation soils. Forensic evaluation of the recovered geogrids allowed determination of the mobilized tensile strains in the geogrid panels along the height of the wall, which revealed a relatively uniform distribution of tension with depth. Figure 4.10. Measured and AASHTO-predicted connection loads in geogrid layers 6 and 10 at the FM load-carrying GMSE bridge abutment.
42 4.3.2 NUMERICAL SIMULATION OUTCOMES THAT SUPPORT THE PROPOSED RECOMMENDATIONS The numerical simulations conducted as part of the NCHRP Project 24-41 study (Section 7 of this NCHRP Project 24-41 Final Report) provide additional justification to the revisions to AASHTO design approach presented in Chapter 3.3. They also provide insight on relevant differences between the response of GRS piers and GRS-IBS structures. Based on numerical simulations of GRS piers, the connection loads T0 was reasonably smaller than the Tmax values, as the geotextile reinforcements were frictionally connected to the facing blocks. The predicted Tmax value was found to develop at the center of the pier, where the surcharge-induced vertical stresses were the highest. However, T0 developed as a connection force at the interface between the geotextile reinforcements and the CMU blocks. In fact, the reinforcement spacing was found not to have a significant effect on the predicted connection force T0, independent on the type of reinforcement or magnitude of reinforcement stiffness adopted in the analyses. Figure 4.11 illustrates this response for the cases of analyses conducted with reinforcements of the same stiffness but varying vertical spacing. In addition, Figure 4.12 shows the distribution of connection loads predicted in mini-piers when using the same ratio of reinforcement stiffness to reinforcement vertical spacing. In both cases, a reasonably uniform connection load was predicted with depth (except for loads predicted the base of the mini-piers).
43 Figure 4.11. Distribution of connection loads predicted in mini-piers for different reinforcement vertical spacing using the same reinforcement type. Figure 4.12. Distribution of connection loads predicted in mini-piers when using the same ratio of reinforcement stiffness to reinforcement vertical spacing. While the numerical simulations of mini-piers showed a significant difference between the predicted Tmax and T0 values, the numerical simulation of GRS-IBS showed that the location of Tmax was very close to the connection between geotextile and CMU facing blocks. Consequently, the connection load was found to be similar to the maximum tension in the reinforcement, justifying the recommendation of adopting Î» =1 in Equation 3.5. This is illustrated in Figure 4.13 for conditions corresponding to self-weight loading and of to bridge loading acting on the GRS- IBS system. Figure 4.14 shows a similar response but for simulations conducted using structures that involved different reinforcement vertical spacing while maintaining the same stiffness to vertical spacing ratio.
44 (a) Under self-weight of the wall (b) Under bridge slab and traffic load Figure 4.13. Effect of reinforcement spacing on the connection force of the reinforcement: (a) Under self-weight of the wall; (b) Under bridge slab and traffic load. 0 1 2 3 4 5 6 0 0.5 1 1.5 2 D ep th (m ) T0/Tmax 0.2 0.4 0.6 Spacing (m) 0 1 2 3 4 5 6 0 0.5 1 1.5 2 D ep th (m ) T0/Tmax 0.2 0.4 0.6 Spacing (m)
45 (a) Under self-weight of the wall (b) Under bridge slab and traffic load Figure 4.14. Effect of combination of reinforcement stiffness and spacing on the connection force of the reinforcement: (a) Under self-weight of the wall; (b) Under bridge slab and traffic load. 4.4 EFFECT OF SV ON STRESS DISTRIBUTION AND DESIGN OF BEARING SEATS In order to provide the basis for the recommended effect of Sv on stress distribution and the design of bearing seats, as presented in Chapter 3.4, the different research outcomes from this NCHRP study were holistically assessed. They are presented next. 4.4.1 FIELD RESEARCH OUTCOMES THAT SUPPORT THE PROPOSED RECOMMENDATIONS For the case of geosynthetic-reinforced soil structures with closely-spaced-reinforcement, the results from field monitoring programs and numerical simulations show that the maximum vertical stress induced at various depths below the bearing seat agrees reasonably well with predictions obtained from analytical solutions such as that of Boussinesq. However, the maximum vertical stress was found to also agree well with the predictions obtained using approximate solutions such as pyramid stress distribution, especially when considering its truncation as recommended by Berg et al. (2009). 0 1 2 3 4 5 6 0 0.5 1 1.5 2 D ep th (m ) T0/Tmax Tf = 70 kN/m, J = 700 kN/m, Sv = 0.4 m Tf = 35 kN/m, J = 350 kN/m, Sv = 0.2 m 0 1 2 3 4 5 6 0 0.5 1 1.5 2 D ep th (m ) T0/Tmax Tf = 70 kN/m, J = 700 kN/m, Sv = 0.4 m Tf = 35 kN/m, J = 350 kN/m, Sv = 0.2 m
46 Figure 4.15 shows the stress distribution before and after placement of the bridge superstructure at the FM load-carrying GMSE bridge abutment. It should be noted that the dashed lines in the figure correspond to measured values while solid lines correspond to predictions obtained using the Boussinesq approach. As illustrated in the figure, there is good agreement between measured vertical stresses and analytically predicted values. Figure 4.15. Stress distribution within the GMSE mass during the various construction stages of the FM load-carrying GMSE bridge abutment (Note: stress values in kPa). The effect of the bearing bed zone on the vertical stress distribution within the fill of the Virginia GRS-IBS structure was evaluated based on the stress distribution within the structure after placing the bridge slab load. The evaluation was conducted by comparing the field monitoring data from EPC with the stress distribution estimated using AASHTOâs 2:1 approach and FHWAâs recommended Boussinesq theoretical approach, as used in design of MSE wall and GRS-IBS. Figures 4.16 and 4.17 provide the results for abutments A and B, respectively, of the GRS-IBS constructed in Virginia and reported in detail in Section 6 of the NCHRP Project 24-41 Final Report. Overall, the stress distribution profile obtained using field monitoring results was in agreement with the predicted stress distributions. It was observed that the load from the slab was transferred to the foundation through the bearing bed zone. The thickness of the bearing
47 bed was the same in both abutments, with the bearing bed proving effective in reducing the applied slab load to about half. However, this reduction was more evident in the case of Abutment A than in Abutment B. This is most likely because the vertical stresses in Abutment A (with smaller beam seat width) were higher than those in Abutment B. Figure 4.16. Vertical stress distribution due to slab load in abutment A of VDOT GRS-IBS. Note: Field data refers to pressures measured from the EPC shown in the insert of this figure (i.e. EPCs 1, 4, 7, and 10).
48 Figure 4.17. Vertical stress distribution due to slab load in abutment B of VDOT GRS-IBS. Note: Field data refers to pressures measured from the EPC shown in the insert of this figure (i.e. EPCs 1, 4, 7, and 9). 4.4.2 NUMERICAL SIMULATION OUTCOMES THAT SUPPORT THE PROPOSED RECOMMENDATIONS The results of numerical simulations of mini-piers revealed a significant effect of the reinforcement spacing on the profile with depth of additional vertical stresses induced by surcharge loads. However, the results of numerical simulations of GRS-IBS structures showed a comparatively smaller effect. Figure 4.18 shows the effect of the reinforcement spacing on the distribution with depth of surcharge-induced vertical stress Îp at 0.5 m (1.6 ft) from the back of CMU facing blocks. It also shows the predicted surcharge-induced vertical stress predicted using the AASHTO 2 to 1 truncated method. The results shown in the figure illustrate that the surcharge-induced vertical stress Îp was not significantly affected by the reinforcement vertical spacing. An increasing reinforcement spacing was found to result in a slightly lower surcharge- induced vertical stress. AASHTOâs 2 to 1 truncated method was found to predict the surcharge- induced stresses that were in reasonably good agreement with the numerical predictions.
49 (a) Induced by bridge slab weight (b) Induced by bridge slab weight and traffic loads Figure 4.18. Effect of the reinforcement spacing on the surcharge-induced vertical stresses: (a) Induced by the weight of the bridge slab; (b) Induced by the weight of the bridge slab and the traffic load. Figure 4.19 shows the combined effect of reinforcement stiffness and vertical spacing on the distribution with depth of surcharge-induced vertical stress Îp applied at 0.5 m (1.6 ft) from the back of CMU facing blocks. The results show that the surcharge-induced vertical stress was similar for the two cases being compared, except for comparatively minor differences towards the top of the wall. Also in this case, AASHTOâs 2 to 1 truncated method was found to lead to surcharge- induced vertical stresses that are in reasonably good agreement with the numerical results. 0 1 2 3 4 5 6 0 30 60 90 120 D ep th (m ) Additional vertical stress Îp(kPa) 0.2 0.4 0.6 AASHTO 2To1 Trunctaed Spacing Sv (m) 0 1 2 3 4 5 6 0 50 100 150 200 D ep th (m ) Additional vertical stress Îp(kPa) 0.2 0.4 0.6 AASHTO 2To1 Trunctaed Spacing Sv (m) AASHTO 2To1 Trunctaed plus surcharge q
50 (a) Induced by bridge slab weight (b) Induced by bridge slab weight and traffic loads Figure 4.19. Combined effect of reinforcement stiffness and vertical spacing on surcharge-induced vertical stresses: (a) Induced by the weight of the bridge slab; (b) Induced by the weight of the bridge slab and the traffic load. 4.5 EFFECT OF SV ON VERTICAL AND LATERAL DEFORMATIONS Reinforcement vertical spacing was found to significantly affect the vertical and lateral deformation response of GMSE structures, particularly under bridge loading. This effect was identified as a result of the parametric numerical evaluation reported in Section 7 of the NCHRP Project 24-41 Final Report. The lateral deformation limits reported in the two current national design guidelines, AASHTO (2017)/FHWA 2009 (Christopher et al. 1990) for GMSE structures and FHWA (2012, 2018) for GRS-IBS were compared to these numerical predictions as well as to data from field monitoring projects reviewed in the NCHRP Project 24-41study in order to establish recommendations for revising the AASHTO design specifications. The AASHTO (2017)/FHWA (2009) method provides a first order, empirically derived approach to evaluate maximum anticipated movement during construction. The approach uses the information presented in Figure 4.20 to define the reinforced soil wall deformation coefficient (Î´R) for a given reinforcement length (L) to wall height (H) ratio. 0 1 2 3 4 5 6 0 30 60 90 120 D ep th (m ) Additional vertical stress Îp(kPa) J=1050kN/m, Sv=0.6m J=700kN/m, Sv=0.4m J=350kN/m, Sv=0.2m AASHTO 2To1 Trunctaed 0 1 2 3 4 5 6 0 50 100 150 200 D ep th (m ) Additional vertical stress Îp(kPa) J=1050kN/m, Sv=0.6m J=700kN/m, Sv=0.4m J=350kN/m, Sv=0.2m AASHTO 2To1 Trunctaed AASHTO 2To1 Trunctaed plus surcharge q
51 Figure 4.20. Variation of the reinforced soil wall deformation coefficient (Î´R ) with L/H ratio. The curve in Figure 4.20 can also be expressed as follows: = 11.81 â 42.25 + 57.16 â 35.45 + 9.471 [Equation.4.3] Using the calculated value of Î´R, the maximum lateral deformation can then be calculated for extensible geosynthetic type and inextensible steel type reinforcements, as follows: = ( ) [Equation 4.4] = ( ) [Equation 4.5] where: Î´R is a dimensionless reinforced soil wall deformation coefficient, L is the reinforcement length (in units of H), H is the wall height measured from the top of the leveling pad (in units of either ft. or m), and Î´max is the maximum lateral deformation of reinforced soil wall (in units of H). The basis for having two forms of the equation (i.e. for inextensible and for extensible) pertains to the global stiffness of the wall (Sr) (i.e., the average reinforcement stiffness over the wall face
52 area). The prediction equation uses the L/H ratio (ranging between 0.3 and 1.175) and the reinforcement type. It should be noted that this equation was developed to predict the maximum lateral displacement of a wall during its construction, which is rarely measured in practice. While this model does not directly predict the additional lateral displacement that can take place upon the application of surcharge loads, Christopher et al. (1990) stated that for a 6-m (20-ft) high wall, each additional 20 kPa (417 psf) of surcharge load results in a 25% increase in the relative deformation. The FHWA (2012,2018) method of evaluating lateral deformation relies on vertical deformation data collected from mini-pier tests, and established the following approach to determine the lateral deformation DL: = 2 , [Equation 4.6] where: DL is the maximum lateral displacement, bq,vol is the width of the load along the top of the wall including the setback, and DV is the vertical settlement of the load, and H is the height of the structure. = , = = 2 [Equation 4.7] where: ÎµL is the maximum lateral strain, and ÎµV is the vertical strain. Both methods have been previously compared to field measurements along with a number of other methods proposed in the technical literature. Bathurst et al. (2010) reviewed vertical limits established by 11 design codes as compared to a database of wall performance. In that study, they concluded that the AASHTO/FHWA method provided a reasonable upper limit in most cases for end-of-construction movements for walls constructed on firm foundations. However, both the vertical spacing of the reinforcement Sv and the facing varied considerably in the walls evaluated. The study also evaluated a careful set of full-scale wall tests that revealed that end-of-construction deformations are influenced by both compaction effort and global reinforcement stiffness when other factors remain unchanged. This effort is somewhat related to changing the spacing of reinforcements with the same modulus. In this case, the wall global stiffness would increase proportionally to the decreased spacing, as reviewed later in this section. Based on a comparison of seven methods, a study by Khosrojerdi et al. (2017) the method reported in FHWA (2012) was found to be one of the best in predicting lateral deformations. Specifically, this method was only found to be slightly unconservative, as it predicted lateral deformations that were on average 88% of the actual measured values. The method was also found to have good reliability when compared to other methods with a low COV value of 0.51 obtained for the 12 structures evaluated using that method. As noted by the authors, a limitation of this method is that the magnitude of the structureâs vertical settlement must be known to
53 predict the lateral deformations. Otherwise, estimates of vertical settlement would need to be made, which would potentially increase the uncertainty of this method. Khosrojerdi et al. (2017) also evaluated the AASHTO/FHWA method which, although it was found to be highly conservative, it was reported as having a fair reliability. The authors noted that the AASHTO/FHWA method requires few input parameters (height of the wall, length of reinforcement) and, consequently, can be used to provide a rough estimation of the maximum lateral deformation of GMSE walls even before construction. The authors also contribute the conservatism to different design and construction conditions before 1990, upon which this method was based, however, the Bathurst results were based on some walls constructed after that period of time. The conservatism that they identified in the method is likely due to the use of highly select backfill for the test walls that were the basis for the Khorsrojerdi et al. (2017) study. As indicated in their paper, âwhich may lead to smaller displacements compared to typical walls in the field because of special attentions in the test wall constructions.â To overcome this apparent bias in the evaluation of the AASHTO/FHWA method, experimental GMSE wall structures with reliable lateral deformation data were carefully selected as part of the work conducted in the NCHRP Project 24-41 study to calibrate the AASHTO/FHWA prediction method. The walls include the following: â¢ Two walls constructed in Stockbridge, Georgia in 1994 (Ling and Leshchinsky 1996). These walls are the same structures evaluated in Section 4.3 of the NCHRP Project 24-41 Final Report (University of Delaware Studies). These walls were constructed with an L/H ratio of 0.3. While this L/H value is much smaller than that specified in AASHTO, it is still in the L/H range for which the AASHTO/FHWA lateral deformation prediction model was developed. The lateral deformations of these walls were monitored during construction as well as after construction completion and application of fill surcharge. The vertical reinforcement spacing values in these walls were 0.4 and 0.8 m (1.3 and 2.6 ft). These structures were selected because of the availability to the authors of the displacement data collected during construction, which is deemed a rare dataset. â¢ One wall reported by Begnini et al. (1996). This wall was monitored after construction, so it provided data for calibration of surcharge-induced displacements only. This wall was constructed with an L/H ratio of 0.34 and vertical reinforcement spacing of 0.5 m (20 in.). Additional displacement readings were recorded with time in this wall under constant surcharge after the maximum load was applied. Only displacement data due to increasing surcharge load was collected. This structure was selected to account for surcharge- induced lateral displacements (i.e. post-construction lateral displacements). â¢ One wall reported by Bathurst et al. (1993), which was monitored after construction and, consequently, it provided data for calibration of surcharge-induced displacements only. This wall was constructed with an L/H ratio of 0.70 and vertical reinforcement spacing of 0.8 m (32 in.). Additional displacement readings were recorded in this wall with time and constant surcharge after maximum surcharge was reached. Only displacement data due to increasing surcharge load was collected. This structure was selected to account for surcharge-induced lateral displacements (i.e. post-construction lateral displacements).
54 The AASHTO/FHWA model was modified to account for the global stiffness of the structure, which was the genesis of the model development . The proposed modifications include the following: (1) Modifying the equation to become a function of the J/Sv ratio; and (2) Introducing a surcharge-induced component into the equation. The resulting formulation is as follows: = 11.81 â 42.25 + 57.16 â 35.45 + 9.471 [Equation 4.8] = . 1 + 1.25 ( ) [Equation 4.9] where: Î´R is an empirically derived relative displacement coefficient (dimensionless), H is the height of the wall and the units of H in equation 4.9 defines the units of the maximum estimated lateral deformation Î´max, J is the reinforcement tensile stiffness defined by the secant modulus at 2% strain, Sv is the reinforcement vertical spacing (in units of H), q is the surcharge magnitude, and po is atmospheric pressure introduced in the equation for normalization purposes. It should be noted that the second term (multiplier) in Equation 4.6 includes in turn two terms: the unity term, which corresponds the maximum lateral displacement during construction, and the surcharge term, which corresponds the maximum lateral displacement induced by surcharge only. Figure 4.21 presents the predicted versus measured maximum displacements for the four structures evaluated in the NCHRP Project 24-41 study. For those walls whose displacements were only measured post construction (including surcharge-induced displacements), Equation 4.9 was used but without the unity term.
55 (a) (b) Figure 4.21. Predicted versus measured maximum displacements using the proposed model: (a) Modified FHWA Method; (b)FHWA 2018 Method.
56 The comparison shown in Figure 4.2 reveals good agreement between the lateral displacements predicted using the proposed modified method and field displacement measurements. Consequently, the modifications proposed for the equations are deemed adequate for incorporation into AASHTO for prediction of lateral displacements in GMSE structures with closely-spaced reinforcements. Based on these findings, it is recommended that in cases where the vertical displacements are known (can be accurately predicted) in load-carrying GMSE bridge abutments, the lateral deformation be estimated as approximately 2 times the vertical movement, which is in agreement with the method in FHWA (2018). However, in cases where vertical displacement measurements are not available, the proposed modified AASHTO approach be used for both free standing walls as load-carrying GMSE bridge abutments (or walls with generic surcharge loads). The recommended approach should only be adopted for the case of walls constructed using high quality backfill, which were the basis for calibration of the model. 4.6 EFFECT OF SV ON BUMP AT THE END OF THE BRIDGE In order to provide the basis for the recommended effect of Sv on the development of a âBump at the End of the Bridge,â as presented in Chapter 3.6, the different research outcomes from this research project were holistically assessed. The recommendations apply for the case of load- carrying GMSE bridge abutments, which as a design concept already aim at minimizing the bump at the end of the bridge. Among these structures, systems designed with closely-spaced reinforcement (e.g. GRS-IBS) were analyzed. The basis for the recommendations to the AASHTO Design Method are presented next. 18.104.22.168 FIELD RESEARCH OUTCOMES THAT SUPPORT THE PROPOSED RECOMMENDATIONS The potential reduction of the bump at the end of the bridge when adopting closely-spaced reinforcement was evaluated using field data reported by others as well as site visits to three projects conducted as part of the NCHRP Project 24-41 study. The structures whose data was gathered during the NCHRP Project 24-41 study include the VDOT wall project (reported in Chapter 5), the FM CDOT project (reported in Sections 4.5 and 4.6.4 of the NCHRP Project 24-41 Final Report), and the MnDOT GRS-IBS project, also referred to as Rock County GRS-IBS (Reported in Section 4.6.2 of the NCHRP Project 24-41 Final Report). At each of these sites, the existence of a bump at the bridge superstructure was monitored through visual inspections, level monitoring of the road surface during site visits, and survey data available from the relevant state DOT. The condition surveys during field visits focused mainly on the intersection of the bridge slab and the integrated approach where a potential bump is expected to be visible. At the VDOT wall project, no bump was observed at the intersection of the bridge slab and the integrated approach on the road surface, as shown in Figure 4.22. Abu-Hejleh et al. (2002) reported on the transition from bridge deck to approaching roadway at the FM load-carrying GMSE abutment and reported no indication of the development of the bump at the end of the bridge. Complementing this initial reporting, the results in Figure 4.23 shows the results of surveys of the FM bridge as
57 collected during the initial 5 years after opening to traffic and as recently as 2016 (i.e. 17 years after opening to traffic). As shown in figure, the profiles reveal no development of a âbump at the end of the bridgeâ. On the other hand, the visit to the MnDOT GRS-IBS project revealed that âbumpsâ had developed at both ends of the bridge, as shown in Figure 4.24. Survey data provided by MnDOT provides a direct measure of the magnitude of the bumps as shown in Figure 4.25. It should be noted that this structure had a significant slope and skew, which may have contributed to the occurrence of the observed nonconformities. Figure 4.22. Road surface level monitoring at the intersection of bridge superstructure and integrated approach 2.5 years after construction in VDOT GRS-IBS. Figure 4.23. Profilometer measurements at the Founders/Meadows bridge abutment.
58 (a) (b) Figure 4.24. Settlement at the MnDOT GRS-IBS: (a) North Abutment; (b) South Abutment. Figure 4.25. Longitudinal Profile at the North and South ends of the MnDOT GRS-IBS project. In addition to site visits, a report providing IRI data and surface profiles of the approaching roadways over five GRS-IBS structures constructed in St. Lawrence, New York were reviewed. The projects were constructed from 2009 to 2011. The surveys were performed several years after construction (2013) and the sites were visually observed in 2013 as part of this research project. Neither the visual observations nor the profiles showed a âbumpâ at the transition from the approaching roadways and any of the bridges. 4.6.2 NUMERICAL SIMULATION OUTCOMES THAT SUPPORT THE PROPOSED RECOMMENDATIONS Considering the results from field observations, especially the development of a bump at the ends of the MnDOT GRS-IBS, numerical simulations were conducted to evaluate the potential occurrence of a bump. Based on the numerical predictions of a structure representing a generic
59 GRS-IBS, the potential for the development of differential settlements at the end of the bridge slab could be identified, as shown in Figure 8.4.26. However, a decrease in reinforcement vertical spacing was found to have a significant impact in reducing the bump at the end of the bridge slab. Figure 4.26. Effect of reinforcement spacing on the bum at the end of the bridge due to traffic load. Figure 4.27 shows the combined effect of reinforcement stiffness and spacing on the development, due to traffic loads, of a bump at the end of the bridge. When the same ratio J/Sv (of Tf/Sv) is maintained, the impact of reinforcement vertical spacing was found to be comparatively small. However, it should be noted that the pavement structure of the approaching roadway was not modeled in this simulation, which could be credited with reduction in the potential development of a bump (FHWA 2012, 2018). Overall, the numerical predictions indicate significant benefit of increasing the reinforcement density under the bearing seat or bridge wing walls, beyond the reinforcement required to provide the target total tensile capacity. 0 5 10 15 20 25 0 1 2 3 4 5 6 7 8 9 10 11 A dd iti on al se ttl em en t ( m m ) Distance to the center of the bridge slab (mm) 0.2 0.4 0.6 Spacing (m) Bump at the end of the bridge
60 Figure 4.27. Combined effect of reinforcement stiffness and spacing on the bump at the end of the bridge due to traffic load. 0 4 8 12 16 20 24 0 1 2 3 4 5 6 7 8 9 10 11 A dd iti on al se ttl em en t ( m m ) Distance to the center of bridge slab (m) J=1050kN/m, Sv=0.6m J=700kN/m, Sv=0.4m J=350kN/m, Sv=0.2m Bump at the end of the bridge